Steel stress for cracking design at serviceability state in RC members

[tkd77]

Hi All,
Some advice needed - primarily to confirm if I am overthinking something I have never worried about before. I work for a large precast producer, who often makes very long, slender columns.

Two considerations often govern the unit checks that we perform.
A) We often have to specifically check for the pitching/tilt lifting case (we operate in the architectural market, so nearly always have to pitch lift from the end (top as installed) of the column as clients do not want lifter recesses on the side face. Whilst the ultimate limit state design is nearly always satisfied by the permanent works rebar specification, we often end up increasing main bar dia and/or overall steel area to limit crack widths.
The query here is whether it is correct to assume an average stress in the reinforcement for the crack checks (i.e. to calculate the tensile stress by a simple force/bar area check). As the main bars are typically different diameters (i.e. we might have B25 in the corners, but B16 in the centre of the face) I often wonder whether we should be proportioning the stress at serviceability limit stage according to the relative axial stiffness of the bar (i.e. using AE/L for the specific bar - which typically cancels to just the A). Thoughts please.

B)Often these are double storey or triple storey columns, and often have a physical void at the floor levels (e.g. a double height column would effectively be two smaller columns but produced as a single member with the main reinforcement running through the void. This section of the reinforcement is designed to prevent a combination of compression buckling and local moment at ultimate limit state - both for the pitching case, and the case when installed when it effectively takes the compression load from the upper sections of the column.
Prior to me taking over, standard practice in calculating the compression buckling capacity was to assume an effective length of 0.7 x the bar length between the concrete sections. I have since changed this to 1.5x the bar length between the end links in the concrete sections during pitching(on the basis that this section effectively behaves as in a semi-sway or sway type mode - i.e. like a truss where the central bay works on vierendeel truss principles) and between 1.5 and 2 for the case when installed (as overall column curvature is that of a cantilever). Am struggling to find any formal or written justification for this method - the only thing I can think of is using a frame effective length coefficient method similar to that found in appendix D and E of BS5950 - but setting the beam values to zero to effectively treat these bars as a compression member of two different section sizes and restraint conditions. Have used this in the past to analyse very old portal frame columns where the column is supporting a gantry crane and formed of two different UB sections. Thoughts on this welcome..... see sketch attached.

 

[Lomarandil]

A) I think strain compatibility would equalize (or effectively so) stresses due to axial tension. I imagine you're also checking the flexural tension during the lift.

B) Agreed that a away factor greater than 1 is appropriate. Something even more than 1.2 per AISC for fixed and fixed(sway), assuming some elastic deformations lead to rotation between the two segments. But I don't have anything more definitive than that short of buckling analysis.

----
just call me Lo.

 

[KootK]

A) I can't comment too intelligently on this aspect of things because I'm not familiar with the British methodology for determining crack widths. In terms of real bar stresses, there is a fair bit going on between cracks where the bars are picking up and shedding tension load and doing that at different rates for the different bar sizes. That said, your computation procedure may well be indexed to bar stresses averaged across cracks as is often the case for evaluations of flexural tension cracking. Whatever the British equivalent of ACI224.2 is, as long as you're running your computations consistently with that, I'm sure that you'll be in good shape.

B) I do not feel that a sway mechanism needs to considered for the case of bar compression induced predominantly by flexure. Any sway mechanism that forms would need to sway-deform all of the bars of the cross section, including the bars in tension. And, for every bar in compression suffering from destabilizing P-delta effects, there will a corresponding bar in tension benefiting from stabilizing P-delta effects. I believe that the two effects will cancel one another out. Naturally, you'll have some degree of shear demand to deal with at the void and that will necessitate treatment of the rebar as a little beam-columns to account for the vierendeel action. That doesn't change the buckling analysis however. Whether this was the logic used by your predecessor, or they just didn't think about sway and got lucky, your guess is as good as mine (probably better).

 

[r13]

A) I think it is appropriate when bars spaced equally (s = constant). Otherwise, use A_TR in computation maybe more realistic.
B) You are quite conservative. See chart below.

 

[tkd77]

Thanks all,
It was a pretty late one when I wrote the query. Woke up this morning and convinced myself that (A) wasn't an issue - both because of strain compatibility and the wording of BS EN 1992 where is discusses average mean strain in the concrete and steel.

On (B) I think I'll stick with the 1.5...but appreciate that sway factors can be lower. Part of the potential conservatism comes from a background in steel design and some of the glaring holes in the old British Standard concerning sway (i.e. it was possible to have non-sway, sway sensitive structures and conversely, non-sway sensitive sway structures...

Arguably we don't get too much of a sway mechanism in the flat lifting case, albeit there is a little from the shears - but we have had a bar buckle on site before (again, as usual everyone on site swears blind it must have been the design..I suspect the installation...but as we manufacture, install and act as the main contractor such honesty tends to result in less than favourable views at the top of the tree. I find sometimes that the ability of structures to not fail when they should is both a blessing and a curse. If I had a penny for every time someone has said "We've never done it like that before...I'd probably have around £8.63

When I get a spare 5 minutes (and if anyone has worked in precast you know that this is a pipe dream), I might model some solid and equivalent truss structures to see what happens.

Thanks again.

 

[Tomfh]

I agree with making it robust. If you can make it less vulnerable to installation mishaps that’s a good thing.

 

[r13]

Actually I'll pay more attention to the device under the hook (lift point), as the failure is more likely to occur. It is a general practice to design the lifting devices with a safety factor of 4 or more, but rarely talked about the device in the structure been lifted. In the past, we used to specify steel or rebar ring in the precast concrete panel for lifting, the practice has since changed after several accidents involving shear failure of the steel during operation. To the bottom line, the reasons caused those accidents were not directly related to strength of the steel, but construction incidents, which are very difficult to manage.