Does tension tie force translate to horizontal shear at the top of column 1

[slickdeals]

The title of my post says it all. Please see attached sketch.

 

[BAretired]

A cold joint introduces a plane of weakness which is not present in a monolithic pour.

There have been a lot of compression tests performed on tied and spirally reinforced columns over the years. The mode of failure does not suggest a need to check shear-friction in monolithic columns.

BA

 

[hokie66]

Does nobody else object to the whole concept of shear friction? To me, it is like a passive rock anchor, in that for the clamping force to be activated, there must be movement/elongation. In a concrete joint, you don't want that to occur.

 

[dcarr82775]

I stand beside KootK in my pro shear friction stance. I am not quite as excited about it as he is, but I still think it is a valid approach. Different topic for a different post.

And yes, there is shear at the joint that needs to be handled.

 

[BAretired]

To some extent, I share hokie's objection to shear-friction in the usual case where the main reinforcement is normal to the cold joint or the angle is in the direction of the shearing force.

In this case, I have a stronger objection. The longitudinal column reinforcement is in the wrong direction to provide shear-friction across the cold joint. They will tend to break out of the column. In this case, I would not be willing to rely on shear-friction.

BA

 

[MrHershey]

@KootK: Maybe I'm misunderstanding things if the opposite is true.

Point out where I'm going wrong here:

24"x24" tied axial member, 4000 psi
Ignore Phi factors for simplicity (the difference is large enough that they won't matter)

Compression strength, ignoring rebar:
0.8*0.85f'c*Ac = 0.8*0.85*4 ksi*24"x24" = 1567 kips

Translate that into a 'shear friction' along a 45 degree plane (or any other plane over 20-25 degrees from normal):
1567 sin 45 = 1108 kips

Calculate out shear friction capacity along that plane using the maximum allowed for shear friction under ACI 318-05 (11.7.5):
Max = 0.2f'c*Ac = 0.2*4 ksi*24"x24"/sin 45 = 652 kips
OR
Max = 800Ac = 800*24"*24"/sin 45 = 652 kips

Results in a major deficit, under 60% of the 'required' if the shear friction mechanism actually applies here. Way larger than the difference in Phi factors (0.65 compressions/0.75 shear = 87%) would suggest.

If my math is correct above, we should be seeing this failure all over the place. But we don't. Why?

 

[KootK]

@ Mark,

I don't see anything wrong with your assessment; it's an interesting case study that I hadn't considered explicitly in the other threads. Here's my critique of your stuff:

1) By choosing to examine a member loaded to it's squash load, you've narrowed the range of members over which your example would be representative. Most members will be subject to considerably less axial load and, as a result, the limits on maximum shear stress will become less important.

2) We're several miles off the reservation here now that we're discussing hypothetical, monolithic shear planes. In this context, I don't feel that the maximum shear stress provisions apply. The limits given in ACI are only there because, beyond those limits, the shear friction equations "may become un-conservative in some cases". It's not as though there isn't more capacity to be had; it's simply not predicted accurately by the design equations.

If you lift the limits on maximum shear stress, ignore the fact that the capacity equations may be bunk, and consider only the clamping force provided by the compression imposed on the column, things work out like this:

P = axial force in column.
V = P x sin(45) = 0.71P = shear on 45 degree plane.
A = V = P x sin(45) = axial force on 45 degree plane.
Vsf = 1.4 x A = 1.4 x V = 1.4 x P x sin(45) = 0.99P = unfactored shear friction capacity.

So Vsf / V = 0.99P/0.71P = 1.40. Always and forever okay. And this doesn't even account for any contribution accruing from the presence of reinforcing steel.

Keep in mind that I truly do not know what I'm doing with this. If you read my other two threads, you'll see that in spades. I've given it a lot of thought, and I have some strong opinions on the matter, but I do not have the answers. I've been inadvertently tagged a shear friction "fan" because I whine about it endlessly. I'm not a big fan of shear friction. Rather, I'm a big fan of trying figure out what heck I'm supposed to be doing with shear friction and why.

The greatest trick that bond stress ever pulled was convincing the world it didn't exist.

 

[KootK]

@BA/Hokie: I think that we're in agreement that shear friction reinforcing in compression is questionable. ACI 318 agrees. R11.6.4.2 reads:

...should only be used when the shear force component parallel to the reinforcement produces tension in the reinforcment...when alpha is greater than 90 degrees, the relative movement of the surfaces tends to compress the bar and equation [whatever] is not valid

.

That being said, a substantial portion of the clamping force in this situation will come from the load itself. And that contribution is 100% reliable because, without the load, there's no shear demand in the joint. That, in part, is why I am comfortable using shear fiction here.

The greatest trick that bond stress ever pulled was convincing the world it didn't exist.

 

[BAretired]

The original construction joint was shown perpendicular to the axis of the column. Our contractor friends thought we were crazy and put it horizontally.

Your contractor friends had no business making such a change without consulting you. If there is any doubt about the capacity of the cold joint to resist shear, remedial measures may be required. Cost to be borne by the contractor.

BA

 

[MrHershey]

Do agree that the shear friction limits provided in 11.7.5 of 318-05 appear to be extremely conservative for members with any significant compression. Until your angle gets pretty extreme, your clamping force is fairly huge. Even before you look at the effects of any longitudinal reinforcement or ties crossing the plane.

There's been a few articles in the ACI journal that note this. ACI 318-08 increased the limits a bit, but still fall short of some of the recommended limits I've seen in journal articles (including one of the articles specifically referenced by the commentary on shear friction limits, which is odd). Additionally, the increases appear to be directed more at increasing the upper limit based on research with high-strength concrete, but don't really address further increases you can get when you've got a significant clamping/compressive force as well.