Cantilevered Beam to Column Connection 1

[southard2]

OK so I designed a beam to column connection with the beam cantilevering over a column cap plate. Its a very small project and normally I might have never even seen the in field condition because the project is so small. But I did and a instantly noticed that the cap plate on the column was missing. They welded the bottom flange of the wide flange beam directly to the wide flange column. This is a braced frame with two column and one beam. On one end the beam bears on top of the column and was modeled as a pinned connection. One the other end the beam cantilevers about 10 feet and I modeled that connection as a full moment connection using four bolts through the beams bottom flange into the column cap plate.

Because the beam is deep and continuous very little moment actual transfers into the column (only like 342 lb-ft). The resulting axial load into the column from the beam is 18 kip. So for the most part this is a typical beam to column connection with just a bit of moment.

Normally I'd tell them to do it over and built it right but this guy is on the city commission. So before I tell him to repair the situation I wanted to explore other analysis options.

My first thought was to treat it as a column to transfer beam connection but again I've always seen a baseplate/cap-plate.

Then I thought I could analysis this as a directly welded moment connection which is commonly used for a W-Beam to W-Column Flange connection. Only in this case the column would be the beam and the beam would be the column. In other words the column in this case is directly welding to the flanges of the beam. Force transfer is force transfer. I could check the beam as if it were the column etc.

Now the typical moment connection usually has the flanged directly welded to handle the moment and a shear plate or angles for the shear transfer. But now that I'm typing this out my thinking may still be flawed because the major load is an axial load which is usualy zero in this type of connection. So while I could easily justify or retrofit the existing connection for shear and moment transfer I'm still in trouble with the axial load.

Has anyone ever seen any design method for a connection like this. All columns I've seen have baseplates and cap plates....so I'm kind of at a loss.

At the bear minimum I'm going to somehow have to explain in layman's terms why has to reshore the structure and add the cap plate. Were talking about a lot of work here. Right now I'd have to say something to the effect of, "It might work, but I can't prove it one way or the other".

Thanks in advance for everyone's thoughts,

John


John Southard, M.S., P.E.

http://www.pdhlibrary.com

 

[hokie66]

At this stage, instead of installing the cap plate which you were expecting, it will be simpler to add web stiffeners to the beam. Line them up with the column flanges. That way, you will have a fully welded and stiffened moment connection.

 

[southard2]

Yeah that is not a bad idea.

I was just reviewing my spreadsheet for the original connection I designed based on the Hollow Structural Sections Connections Manual (Section 5-3 "Continuous Beams over HSS Columns...See page 5-15 for the design example). Yes I know I have a Wide Flange Column but this should be pretty close.

Web Crippling, yielding, and buckling were OK.
The routine also checks for Bending with prying on the flange of the W-shaped beam and Bending with prying on the Cap Plate of the Column. The result for both checks was a cap plate minimum thickness and minimum beam flange thickness of around 0.17 inches. The flange thickness on the W18x35 is 7/16 which more than twice this thickness. This isn't a wind frame so the moment part of this connection is very small (the inertia of the beam is much larger than the column). Most of the moment remains in the beam.

Since web crippling on the beam isn't an issue I think you may be right. Just add web stiffeners and be done with it. At any rate he needs to be punished in some form for such a huge deviation from the plans without even a call to ask if it would be OK. Plus I'm going to make sure he has welded all around and not in just a few areas. Welds are way way good for these loads so I'm not concerned other than the fact that I want to make sure they are there.

Anyone else have an opinion?



John Southard, M.S., P.E.

http://www.pdhlibrary.com

 

[paddingtongreen]

It sounds as if you may not need to do anything. It sounds as though the weld does not see tension. You can take a spreading load path through the root radii and probably find enough stiffened length to handle your forces.

Michael.
Timing has a lot to do with the outcome of a rain dance.

 

[csd72]

If you have checked the capacity of the weld and the web crippling under the worst of the two flange loads then it will be fine.

This is a very first principles type of thing, think of the web over the column as a truss with compression chords over the flanges and a tension diagonal in between then apply the flange forces of the beam and the column to this to obtain what the internal forces are at this joint and then check the capacities according to code.

 

[BAretired]

Add the stiffeners as hokie suggested. It will be a stronger connection than using the cap plate.

BA

 

[southard2]

Thanks everyone for your input. I've decided to add web stiffeners that are as thick at the column flanges below in essence making this a moment connection that has a moment and axial capacity far beyond what I need.

After reading csd72's reply I decided to check the W18x35's web as a axially loaded plate column. It failed by a good bit. Now adding the stiffener's will in essence elimate this problem as the cross section of the web would convert to a W-shape.

What I thought was interesting was that I wouldn't have expected the web to fail as a column since the web crippling checks were all good. And I properly adjusted the equations for the non cap plate condition. Non of the web crippling equations represented a problem. I used the same clear dimension h between flanges in my web as a column analysis too. The web crippling equations are probabably based on testing and may have a lower factor of safety built in. Its also probably that the testing accounted for the fact that the web crippling cross section might increase as you go from bottom flange to the unloaded top flange (arching principles). Not sure why the large descrepancy but I'm going to leave that for another day.

I agree that with the stiffeners the connection will be even stronger than what I had before as it approaches a full moment connection and I know that web won't cripple or buckle then.

Thanks for all the advice. Its really nice that these boards are here. I'm a one man business again and its nice to be able to bounce ideas around. And perhaps more importantly it helps confirm or dispell various lines of reasoning. These message boards have helped me a whole lot in my practice. Even when I'm just quitely reading various postings.



John Southard, M.S., P.E.

http://www.pdhlibrary.com

 

[SAIL3]

Web buckling should not be an issue unless there is a conc load on the bm directly over the col.
Absence of that, the load will be transferred primarily thru shear in the web of the bm and usually local web crippling may be the only issue to be addressed.

 

[a2mfk]

342 ft-lb? Without a sketch or knowing the whole structure, why bother with a moment connection ie why even transfer any moment into the column? Just about anything you would do could transfer this moment...