QSIIN

I have 3 answers for this, two of which were provided by an Earthquake design specialist and one of which I cannot print here or I might get banned (or threatened again).

My simple reply is

"A non-ductile wall in a ductile frame will not last past the second sway in an earthquake, and therefore will not be a gravity resisting vertical element thereafter."

His further more technical addition to this is

"Structural walls tied to the floors and carrying gravity loads form part of the lateral system and must be treated as such. It is incorrect to assume otherwise. The ductility assumptions for all structural walls need to be consistent across the structure unless a displacement based analysis is used to establish the ductility demand on each wall element for the design event."

Most in-situ vertical element within a structure will experience some component of lateral force when subjected to an earthquake (or wind event). Concrete buildings in Australia are likely to have fire stairs and lift cores that act as the primary LFRS. When you perform an analysis in Etabs or similar building analysis package, it is common to reduce the stiffness of slab elements and column elements further than the values given in AS3600 (columns = 0.3 to 0.8 and slabs = 0.4).

In my experience, the common criteria used for engineers to start considering a vertical element as a wall is an aspect ratio of 4:1. Residential buildings tend to have blade columns with vertical element widths generally governed to slot within apartment party walls. I think it is prudent that the engineer considers these as lateral elements for analysis. You may confirm that no additional reinforcement other than the minimum reinforcement required for columns (or wall if forces are small) after considering axial, shear and flexural demands, but this cannot be omitted from the verification process. I'm looking at a model now with a blade columns each with about 1% of the total shear demand on the building from Earthquake. Although small relative to the demand which goes to primary LFRS elements, this shear and flexure must be considered in design.

A site soil classification of D and E prohibits the simplified method of 11.5 for compression force design. Provisions for columns in section 10 must be used for design in this instance including detailing requirement.

Quote (rcassar)

When you perform an analysis in Etabs or similar building analysis package, it is common to reduce the stiffness of slab elements and column elements further than the values given in AS3600 (columns = 0.3 to 0.8 and slabs = 0.4).

Are designers simply using a lower number because they feel like it, or are they doing the required analysis to justify the reduction?

You cannot use less than the cracked inertia, and it is hard to get below .3 on that basis? Technically it should be higher than the cracked value to allow for tension stiffening between the cracks.

Hi Rapt,

I was afraid you'd have an answer along those lines, though thank you for the contributions you can give. It's a shame that's how it is. Any chance I can get that third response privately?

The two you've answered are my impressions as well. With the new amendment coming to AS3600, can we expect some rewording of things for better clarification? I think this is important as, unfortunately, EQ analysis and design is brand new to a lot of very experienced engineers.

Regarding cracking of elements, I would think if the member is uncracked then the analysis should be done using gross member stiffnesses, as the code states. Cracking the columns might be conservative for the Core box design, but if you've got 2-3m long blade walls cracked down so the core does the work, you need to first justify how it's cracked, and second think about the consequences of it actually cracking - does the wall fail? The whole building doesn't need to collapse in an EQ for it to be considered failed.

The code does not teach earthquake design. Designers need to study earthquake design logic before trying to apply it, and that requires good text books. Agent666 or someone from NZ might be able to help with guidance on sources.

The logic is nothing like what people have gotten away with for wind design previously.

There will be a commentary next year, but designers in Australia are traditionally not taught earthquake design and a lot of the more experienced designers still think in terms of "wind design controlling strength so we will just do that and it will cover everything". Designers need to study and understand earthquake design before they read the code!

RE stiffness, if it is uncracked under full earthquake load (not reduced earthquake design load) then yes, it is uncracked for analysis.

The difference in simple terms with earthquake design as required by AS3600 is you
- determine the earthquake load,
- assume a level of ductility which reduces the earthquake load by a factor
- design for strength based on that reduced load
- detail it based on that assumed ductility

So if the assumed ductility does not match the detailed ductility, when the full earthquake load hits the structure does not have the combination of strength + ductility to survive. It you have assumed moderately ductile, then you have designed assuming an earthquake sway of .67 / 3 of the full sway, so .2233 of the full sway. If the structure is not designed and detailed to handle 4.5 times the sway you have designed for, you are in trouble.
If one or more elements in that structure are not designed and detailed for the full sway, they could lose all gravity load capacity. So even if they are not being expected to resist sway, they will no longer be able to resist gravity loads either (hence my simplified comment in my 1st post). You will then be relying on your Robustness design to hold it up. And not many are doing that properly either. For example, AS3600 section 8 deemed to comply beam reinforcing does not guarantee any level of robustness as it does not require continuous bottom reinforcement. Designers have to work it out for themselves.

The next comeback from someone will be "but we are not paid sufficient fees to do all of that". You already know my response to that so I will not say it again.

Even though quite old now, the old Park & Pauley text 'Reinforced Concrete Structures' is a good primer for seismic design (and detailing of concrete structures). These guys were the driving force for adopting the capacity design approaches ingrained in NZ and many international seismic design standards.

But really any half decent text should take you through the basics, I think the issue in Australia is the lack of commentary explaining the 'why' and a lack of people being taught it effectively in universities is a big part of the issue. It's not something new, most Aussie engineers I've worked with over here in NZ have a really hard time picking it up and hence actually being good at it because they weren't even taught the most basic things in university when it comes to ductility and seismic design. You don't know what you don't know..... Stuff that we are force fed from year one at University.....

On a related topic AS3600 14.1(b)(ii) states:



Does this mean the entire building needs to use Class E reo, or is it just the lateral load resisting elements (or somewhere in-between)?

Hi rapt,

Just following on from the above discussion, I understand that the seismic demand on so called "gravity only" blade walls and columns should not be disregarded in design however negligible it may be, but was wondering if we design and detail the main lateral load resisting elements (cores and shear walls) as limited or moderately ductile, do the "gravity only" blade walls and columns need to also be designed and detailed based on the limited/moderately ductile assumption as well, or do they need to be designed as non-ductile based on the full elastic earthquake load?

Does clause 14.4.4.3 - Axial load limit for elements with u > 1 that are limited/moderately ductile play any role in this determination? The clause appears to imply that if the stress under G+0.3Q in any wall/column exceeds 0.2fc', you cannot proceed based on the limited/moderately ductile assumption and will need to design the wall/column element in question as non-ductile for the full elastic earthquake load. What are your thoughts on this?

There is no such thing as a 'gravity only column' in a concrete structure subject to earthquake.

They all go along for the ride from a drift compatibility standpoint. How much curvature a potential yielding region sees in these situations governs how it should be detailed (i.e. what level of ductile detailing is required). I don't have a copy of the 2018 aS standard so cannot answer your direct query, but history tells you assuming gravity only columns is not a wise choice if you want to design for all the possible actions that develop in your columns during an earthquake.

In NZ we went through a period from 1982 to 1995 where some bright spark decided we could classify columns as 'gravity only columns' in our design standard with no consideration of what seismic forces developed as they go along for the ride to the same drift as the main lateral load resisting system. They severely relaxed the detailing requirements with respect to confinement/anti-buckling and shear reinforcement in these sorts of columns (things you generally need to increase the reliability of getting some sufficient ductile response out of your columns). These so called 'non-ductile columns' had some part to play in the two major structures that collapsed in our 2011 Christchurch event, and we've been addressing and strengthening for this condition every since.....

Drapes,

Not sure how you read that into it. There are no tricks that I can see in the wording.

Structural wall definition is in 14.2.10. It is very specific. Basically any wall connected to the floor diaphragms that attract horizontal earthquake or wind action.

The clause is simply controlling the wall dimensions for "All Structural Walls" based on the defined loading. A structural wall, whether it be ductile, limited ductile, non-ductile, if it is structural, the axial load stress is limited to this value.

Thanks Agent666 and rapt for your feedback.

rapt, the clause does not appear to concern non-ductile walls though, as it only relates to elements with a ductility factor of greater than 1.



Also, if this clause relates to all structural walls (including columns), it will be very difficult to satisfy this clause for the majority of columns given that they will be attracting very high axial load at the outset due to gravity (even under the seismic weight given by G+0.3Q).

Finally, as I mentioned before, if I designed and detailed my main lateral load resisting elements (core and shear walls) as limited or moderately ductile and there are no issues on that front, but a handful of columns fail to satisfy the above clause, does that mean my initial assumption of taking limited or moderately ductile becomes invalid, and will need to either increase the size or strength of the columns to suit, or resort to designing the whole structure as non-ductile? As you noted in a previous post, "The ductility assumptions for all structural walls need to be consistent across the structure..."

You may not be recognising that even under mu = 1 global actions, the local ductility demand (i.e. curvature) in certain elements may exceed the limits afforded to ductile detailing limits. I'm just not sure how AS3600 addresses the limits side of things, whether the driver is a curvature limit or otherwise for tipping you into the various detailing requirements. I hope its not based on the global ductility as that mu>1 seems to imply as its an incorrect approach!

This picture from NZS3101 helps explain this concept, that the global ductility for determining the overall loads is independent of the local ductility demand occuring in members. You can for example design for mu=1 for determining the global base shear, but have certain elements that require ductile detailing due to the curvature they are seeing in potential plastic hinge regions.





The reason for the axial load limit is that walls traditionally don't perform that well under higher axial load because the confinement requirement isn't usually as rigourous as for columns.

In NZ for example we have an upper limit on N* of of 0.3 instead of 0.2f'cAg you note. But then traditionally we have always had quite strict confinement criteria when compared to most international standards. If we go over 0.2f'cAg then we need to also ensure the thickness of the wall doesn't make it too slender. Other than that we also have the provision that if you exceed this axial load limit you have the option of designing the wall following the column provisions (which in theory will result in higher levels of confinement).

That makes a lot of sense, however based on my design experience in Australia the global ductility class assumed for the building is used throughout in the design and detailing of the vertical lateral load resisting elements. I am not aware of breaking it up into global and local ductility, and the code certainly does not go into that sort of detail, perhaps due to being located in a region of low seismicity.

Rapt has also alluded to the fact that the "ductility assumptions for all structural walls need to be consistent across the structure..." Further, with reference to a recent paper concerning AS3600 and the design of reinforced concrete walls in regions of lower seismicity, it quotes that the "ductility class assumed for the building should be that of the least ductile element within the lateral load resisting system." Both these statements appear to attest to the fact that a single global ductility should be adopted, at least in regions of lower seismicity like Australia, but I may be missing something?

Notwithstanding the above however, and referring back to clause 14.4.4.3 re: axial load limits, there are no provisions on detailing if you exceed the 0.2f'c stress limit - its a dead end. So if you have assumed a limited or moderately ductile structure, and designed and detailed the main lateral load resisting elements (core and shear walls) to suit, but a handful of columns happen to exceed the 0.2f'c limit (which is highly likely in almost all scenarios as far as I'm concerned), it appears you will need to revert to a non-ductile system for all elements (not a great result), or increase the size or strength of the columns in question in order to maintain the limited or moderately ductile assumption made initially. I would expect (as you noted) that if you exceed this limit, it would simply tip you into more stringent detailing requirements, but there are currently no provisions like that in this regard.

Clause 14.4.4.3 states that the 0.2f'c limit applies to walls; it does not mention columns. I'm not sure if that means as long as you design the wall using Section 10 of the code you don't have to meet this requirement. I suspect that was not the intent of the authors of the code; however, it's certainly not clear.

There does not seem to be a clear definition of the difference between a wall and a column in the code other than 5.6.2.

Drapes,

The logic is meant to get you to increase the column dimensions until the clause is satisfied! It is an absolute maximum, independent of detailing.

Apparently it has been shown that having too high a pre-compression of a wall limits its ability to behave in-elastically and therefore perform its post-elastic plastic drift. A bit like the more stringent ductility limits for beams which have to have sufficient ductility to allow sufficient plastic rotation to occur after a plastic hinge forms.

As you should not be mixing ductility of elements (as you have discussed above) then you cannot simply say the offending wall is non-ductile if all others are limited ductile or moderately ductile.

Thanks rapt, that makes a lot more sense now.

Interesting then that if we choose to proceed based on limited or moderate ductility, that the sizing/strength of the heavily loaded columns (possibly the vast majority of cols on the lower floors) may now be dictated not by ultimate gravity load combinations but by the axial load limit under its seismic weight.

Retrograde, based on discussions herewith I believe "structural walls" under seismic loads relates to all vertical loadbearing elements whether they are walls or columns, that is any wall connected to the floor diaphragm that attracts horizontal earthquake or wind action (that includes columns as far as Im concerned). Remember, if we choose to proceed based on limited or moderate ductility, then we will need to design all walls as columns using section 10 anyway, as the simplified method will not be allowed in this instance as per cl 14.4.4.1.

Drapes

Probably not when you start looking at the numbers. Seismic weight is a lot lower than factored gravity load and gravity load combined with wind will have significant moments associated making the controlling stress under gravity load and possibly wind a lot higher than the seismic load stress which has no moment effects included.

rapt, agree however you can use the longitudinal reinforcement to your advantage under factored gravity load or gravity load combined with wind, but for the axial load limit check under seismic weight the reinforcement is disregarded

Quote:

That makes a lot of sense, however based on my design experience in Australia the global ductility class assumed for the building is used throughout in the design and detailing of the vertical lateral load resisting elements.

Took us 35 years to sort it out (our code up unto 2006 used to be exactly the same), you'll get there eventually to.

But in the interim if that is true with the detailing required being tied to the global ductility then you're certainly doing it incorrectly. It has nothing to do with low seismicity.

Don't forget however that in Australia, many high-rises are making use of "blade columns" designed to Section 11, a Section which might as well not even have slenderness limits in it as they are significantly more relaxed than Section 10 (no moment magnifier for walls..)

This means many buildings have 1000x200 or similar "walls" that are slender and have no confinement, but are considered too small to "attract" significant lateral loads, and are therefore "gravity only".

This is how engineers will continue to get away with these designs.

The only thing in the new code that limits this is the requirement for confinement in walls with 65MPa or more. But as this references Section 14 EQ section, I've already seen designs that say the confinement isn't necessary as it doesn't attract lateral loads..

Small "columns" that don't attract lateral loads...4-1 ratio, therefore wall, therefore Section 11 capacity...no lateral load therefore no confinement...easy peasy

So nothing has changed...

Quote (Jishin1)

are considered too small to "attract" significant lateral loads, and are therefore "gravity only".

But the 2018 code now states that all walls connected to floor diaphragms attract horizontal loads, so if you consider them gravity only you violate the code.

Lots has changed...

Quote (Jishin1)

Section which might as well not even have slenderness limits in it as they are significantly more relaxed than Section 10 (no moment magnifier for walls..)

Just picking up on something else you said, the slenderness limits under section 11 are actually more stringent than section 10 (He/tw < 30 versus Le/r < 120). Also, the accidental eccentricity term in 11.5.3 provides a moment magnifier.

Retrograde,

The code doesn't explicitly state that. That's what this whole thread was initially about.

I think the 14.2.10 definition of a structural Wall is vague.

Is it:
Wall (either load bearing or non-loadbearing) (connected to floor diaphragms) that attracts horizontal earthquake and wind design actions.

Or:
Wall (either load bearing or non-loadbearing) connected to floor diaphragms that attracts horizontal earthquake and wind design actions


Is it the floor diaphragm or the wall that is attracting the horizontal actions? By defining that an element can attract horizontal actions, it must mean that the opposite can also happen. Otherwise why not just define it as "all walls (and columns) connected to diaphragms".
I'm just stating interpretations I've already seen to get around the new EQ restrictions.



Regarding slenderness, Section 10 only requires a minimum moment if L/r<25. Beyond this a moment magnifier is required, which can be in the order of 4-5x the minimum moment, sometimes more.

Section 11 is just H/t<30. Which most typical walls are, and the eccentricities are just minimum values, far less than potential magnifiers from Section 10. Since radius of gyration, r, is 0.3D, more "columns" need magnification beyond min moments than "walls" would.

This means lightly loaded columns of fail due to slenderness, when their equivalent Section 11 capacities pass with flying colours. So either the moment magnifier is far too conservative, or the wall code is unconservative, or both.

Jishin1, 'wall' is singular so 'attracts' refers to that. 'Diaphragms' is plural so 'attract' would be used if that were the intent. Probably a fortunate escape from modern under-use of commas but clear nonetheless to a reader who isn't trying to 'get around' the code.

Why should code writers bother trying to make the code ironclad against such designers? It's a fine line between endangering safety because it's profitable by relying on ambiguity and endangering safety simply because it's profitable.

steveh49,

That's exactly my interpretation of the Grammer of the definition, lack of commas and all. So if that is the correct grammatical interpretation, then the code isn't saying that all walls connected to the diaphragm attract horizontal loads, as is retrograde's interpretation.

Which, again, allows engineers to bypass some of the strict section 14 detailing for "blade columns" that don't "attract" horizontal actions.

I'm not trying to find a loop hole for myself, I'm just pointing out the loop holes practicing engineers have already found and are already exploiting.

Unless I'm missing something, you'd have to provide some specific detailing to isolate the wall from attracting loading. Unless thee main load-resisting elements are infinitely stiff (not possible) or the wall has zero stiffness (not possible), the relative movement of upper and lower floors will cause the wall to 'attract' some load. AKA no such thing as gravity-only as Agent666 said earlier.

Again, the code can't stop designers ignoring it.

Quote (Jishin1)

The code doesn't explicitly state that.

I would agree with you that the wording in places leaves a lot to be desired.

Thanks for all the responses

I guess there's still no real conclusion though, as any element connected to a diaphragm, no matter how small, will attract horizontal loads, even if it's just a fraction.

Strange that the new code doesn't have similar restrictions for the column section though, as by definition they will attract horizontal actions too by being connected to the diaphragm, unless the lateral system is IMRF.

If this is the case, Section 11 can ONLY be used for blade columns, shear walls, and even interconnected core walls if a mu of 1 is adopted.

And if a mu of 2 is used, ALL walls must be proportioned to be at 0.2f'c, even if designed as a column using Section 10??

Quote (QSIIN)

Strange that the new code doesn't have similar restrictions for the column section though

My thoughts on this are that columns are inherently more ductile than walls given the closed ties and cross ties required by 10.7.4. And therefore the restrictions applied to walls are not required.

QSIN

That is the conclusion.

Any wall connected to a diaphragm will attract horizontal loads and must be designed for the loads attracted.

So must columns according to the Drift requirements in 14.4.2. But there is no simplified column design method to rule out so the previous discussions do not apply. I assume columns are not as badly affected by the high pre-compression effect so do not require the .2f'c limit.

The simplified method can only be used for a limited range of walls, based on loading and stress state.

There are other things to consider that some will not realize, or will ignore. e.g. In the analysis, you cannot just assume that all columns/walls are fully cracked and use a single % of I in the analysis for all columns and walls. You have to check the level of cracking in each element. If a wall is fully in compression, Ig is used. If in tension I reduces depending on the stress state. There is not a single % of I for all.

As I keep saying, AS3600 is not a text book on the analysis and design for seismic effects. No-one should be designing to AS3600-2018 seismic rules without first reading some good text books on the analysis and design logic. The whole logic is foreign to anything they have had to deal with before and you should understand the logic before even reading the code. Maybe then designers will understand what the code is trying to achieve and stop trying to find illogical ways around it. It is nothing like designing for wind and cannot be approached like that.

Agent666 might be able to point people in the right direction on good text books.

rapt,

just to clarify, you mentioned earlier that the 0.2fc' axial load limit check does apply to columns - "The logic is meant to get you to increase the column dimensions until the clause is satisfied."

But now you appear to be saying the opposite - "I assume columns are not as badly affected by the high pre-compression effect so do not require the 0.2f'c limit."

Drapes,

Sorry, I meant "walls" back on the 20th. I will edit that post.

Except I cannot edit back that far!

Thanks rapt, all your subsequent comments make a lot more sense now.

So if the axial load limit does not apply to columns designed strictly to section 10, what about for walls designed as columns but with ligs avoided as per cl 11.7.4? Would the axial load limit apply in this case given no ligs will be present? I imagine it would apply in this instance.

Also, with regards to clause 14.4.2 - Inter-storey drift where it mentions "all vertical load-bearing elements shall be designed for the calculated horizontal drift..." (which includes columns), how do you design something for the "horizontal drift"? Sorry I might be reading into it, but is it simply referring to the horizontal forces that the columns will attract under equake loading based on the ductility assumptions? Are there any other considerations or checks (apart from say P-delta effects) that would need to be undertaken when designing for the "horizontal drift" or is it as simple as designing for the horizontal forces that the cols attract?

If that's the conclusion, then why does the code insist on defining something so vaguely? Why not just say "ALL walls and columns MUST comply with the following" ? Implying that some walls attract horizontal actions implies that some don't.

After 10 years, these vague definitions are the best they could come up with?

The code isn't a how-to guide, sure, but the code is pretty explicit, and very near a how-to guide when it comes to slabs for example. Just look at the detailing rules. You can layout Reo without even knowing what you're doing by following the diagram.

Look at the slab design section, or the slab deflection equations. Has anyone even used them since the computer was invented? Pages and pages of simplified slab calculations and tables, explicit minimum requirements for slabs and beams, but for earthquake, which they are clearly trying to emphasise more and bring more attention to, is still so vague.

Maybe the code doesn't consider earthquakes, or the consequence of error, as important as slab detailing. Afterall, I don't remember the last time an ultimate earthquake event occurred in Australia, so it's just a gamble.

Or maybe the code committee just doesn't want to upset the industry too much, lest the contractors get on their backs about it, so they keep it as vague as possible. These new requirements drastically change the way buildings are being built right now. God forbid the day Australia gets an actual earthquake, the fires will just be a blip by comparison.

Drapes,

The design method used to calculate the wall capacity (chapter 10 or 11) does not affect it. Any wall that is partly in tension has to be designed using column design logic or strut/tie if it is short.

ACI apparently suggests a 3:1 ratio to be a wall. Some parts of AS3600 suggest 4:1 (Fire). Personally I do not like definitions like that as there is not a sudden step in the logic, a 799*400 column is a column while a 800*200 column is a wall!

In a multi-storey building, if you had some very stiff walls, then 4:1 blade columns would probably act more like columns than walls in this regard as the very stiff wall would be in single curvature and brace the relatively very un-stiff walls which would be in double curvature.

If you only had all 4:1 blade columns and no stiffer walls, then they would more likely act like walls in this regard.

I will see if we can get more discussion into the commentary.

Jishin1
Maybe you are too young to have experienced it. Newcastle in 1989, 13 dead, 160 injured and I felt it very strongly in Sydney. Sufficiently strong for people to get out of the building very quickly!

I will not dignify the remainder of your comments with replies!

And it's only taken 31 years to get an earthquake Section that is obviously still too vague for readers to draw clear conclusions.

And so engineers will continue to ignore earthquake.

There has been an earthquake section in AS3600 for the last 31 years. It was not very good as it was created from the ACI logic in the early 1980's and never upgraded as earthquake logic evolved.

The problem before about 1993 was that there were limited areas where earthquake had to be considered, mainly some areas around Adelaide and south of the Harbour in Sydney, controlled by AS1170.4, not AS3600.

Seeing Newcastle is north of the harbour, this was changed after the earthquake with a new version of AS1170.4 in 1993 (ps it had nothing to do with AS3600!).

But unfortunately most Australian designers assumed Wind Loads were worse so they did not have to worry about earthquake anyway.

Other than that hopefully most engineers in Australia will attempt to learn more about earthquake design and once they understand the logic apply the "vague" rules intelligently and in the intent in which they were written.

It seems like (based on the replies of many in this post and other earlier posts) that in Australia there's a persistent base of engineers simply looking for a way to bend the rules/interpret things in a way that is more like what the've been doing incorrectly for the last 31 odd years... rather than simply getting on with it and giving the design and detailing for earthquake forces the credence it deserves.

Honestly it's the best chance you'll get to kill as many people as possible in your career.... go for gold.

Unfortunately, I think there is too much pressure from contractors to have the lowest rebar rate.

The train of thought is that if companies follow the correct detailing rules, they'll not be competitive and lose the job because they're too heavy.

Well, imagine there's been a significant seismic event in Australia sometime in the future and for the sake of a lack of few additional stirrups you know you were required by the codes of the time, you now have to personally live with being responsible for the deaths of 100+ people in a structure you designed.

For some practicing engineers in NZ, this is there reality. Knowing their flouting of the rules directly contributed to the deaths of 100+ people.

I find the lack of accountability and some formal regime for the peer review of projects in Australia quite interesting, as surely once the shoes on the other foot and you as peer reviewer are being asked to put your signature on a design you didn't do but know it doesn't comply would raise some concerns for most engineers wouldn't it. I'd hazard in this situation given the risks most engineers wouldn't feel comfortable signing off the others design as a reviewer, yet as a designer you're quite comfortable flouting the rules in the first place.

In NZ unless you can demonstrate compliance with the standards in plsce during the peer review process, you'll struggle to do your 'own thing' that significantly departs from the code writers intent. You can do alternative things like following other international standards or guidance, but you'd generally only do this if your standard lacked guidance in that area or had recognised flaws or shortfalls, or didn't reflect current knowledge. Codes are a minimum standard to achieve a recognised level of life safety. While we argue about numbers, don't forget what we do is primarily about maintaining life safety in the unfortunate event we see an ULS event.

It seems like Australian concrete standard has made a large leap recently in keeping up with current international best practice, but its outpaced the current knowledge and comfort zone of the engineer's as they struggle to make sense of all this new (to them at least) seismic stuff. As rapts noted, any modern text book should be able to clue people up on the basics. All these rules have a basis in terms of ensuring buildings stand up to real world events.

Self-certification allows shonks to flourish. It isn't a good match to price competition.

OTOH, Engineers Australia has an oversubscribed webinar coming up so plenty want to learn.

Regarding guides, texts, etc. I came across this guide from the SRIA which seemed to cover the basics presented in a fairly straightforward easily understood way with respect to detailing for seismic requirements.

https://www.sria.com.au/pdfs/SRIA_Guide_to_Seismic...

Thanks Agent666, a great resource indeed.

I came across another paper (which I alluded to in an earlier post) which appears to be a precursor to the new earthquake provisions of the code and is titled "RC walls in Australia: seismic design and detailing to AS1170.4 and AS3600".

Rapt, with reference to this paper and going back again to the clause on the axial load limit for structural walls, recall you left off by assuming this limit probably did not apply to columns. Please see below excerpt from the above paper providing some commentary on this, where it initially suggests the axial load limit applies to BOTH walls and columns, however its concluding statement only references walls - I have highlighted these sections for clarity. What are your thoughts on this? Do you still believe this limit would only apply to walls given there will be a higher level of confinement provided by default in columns?

Details of the paper quoted by Drapes:

To cite this article: Scott J. Menegon, John L. Wilson, Nelson T. K. Lam & Peter McBean (2018)
RC walls in Australia: seismic design and detailing to AS 1170.4 and AS 3600, Australian Journal of
Structural Engineering, 19:1, 67-84, DOI: 10.1080/13287982.2017.1410309

To link to this article: https://doi.org/10.1080/13287982.2017.1410309

The paper is available for free download to members of Engineers Australia.

Doug Jenkins
Interactive Design Services
http://newtonexcelbach.wordpress.com/

Drapes,

I discussed this with the author of the paper mentioned above, Scott Menegon, before I gave that reply on Jan 24th.

If a column was the dominate stiffness in a sway frame and in single curvature, it may apply to it. If the column were a minor stiffness braced by more dominate walls and in double curvature, it probably does not apply to it.

Great, thanks rapt

"If the column were a minor stiffness braced by more dominate walls and in double curvature, it probably does not apply to it."

Does this mean then that there is a point where an elements stiffness can be deemed negligible and the tight EQ requirements wouldn't apply? This is essentially what I've been trying to get some clarity on.

If a long blade column is significantly less stiff than the core box, do the EQ requirements for structural walls still apply?

The EA presentation a few nights back, the presentor was asked about different Mu and Sp values for different elements in the same structure (short stocky walls in a high-rise). The presenter's response was that the ductility of a single element doesn't necessarily dictate the ductility of the whole structure, and different values could be adopted for different stages of analysis. This is kind of in conflict to the text under table 14.3

QSIN,

Quote (QSIN)

If a long blade column is significantly less stiff than the core box, do the EQ requirements for structural walls still apply?

If designing in accordance with AS3600's and general Australian simplified earthquake design logic, the answer is still YES.

RE mu and Sp values,
The places where mu = 1 for a specific calculation are there for specific purposes.
e.g. A wall may be fully in compression under mu = 3, so a designer may think he can detail it as a compression only wall.
But under the extra plastic sway (drift difference between mu = 1 and 3), the wall then goes into tension, it must be detailed as a wall with tension. Otherwise, when it goes into tension,it is not designed/detailed to provide adequate ductility in that case.

So AS3600 tells you that the decision as to which design and detailing rules to use is based on Mu = 1, not mu = 3, (PS I have left Sp out of this but it varies as well) so that it is detailed for the failure condition not a simplified "elastic" (elastic in earthquake design terminology) strength condition.

QSIN,

Further to this, my earlier comment to Drapes only applied to the .2f'c limit in 14.4.4.3!

Thanks RAPT,

Cl 14.4.4.3 does state "All structural walls", and by definition (and what this thread is all about) shouldn't the 0.2f'c apply to all elements, regardless of their relative stiffness?

On another note, if the dominant element (say lift core) is in tension with a Mu of 2, it can be designed and detailed to Section 14, with enough tension steel required for the Mu=2 reduced tension force.

But if a wall is in tension with Mu=1 but in compression with Mu=2, what is the tension force to design for? Somewhere in the middle, or just the full Mu=1 tension force?

QSIIN, your last point is precisely what I am still a little confused about as well.

It appears the decision on whether a wall is in tension or compression, or whether the wall will require ligs, and other similar checks, needs to be based on the non ductile loads with mu=1 (full elastic earthquake load), irrespective of the system ductility assumed in the design even if its limited or moderately ductile with mu>1. The checks based on mu=1 will then inform the detailing requirements only (for example if you need vert bars on each face in lieu of central, or if ligs are required), however as far as the overall strength design is concerned this can still be based on the reduced loads with mu>1 provided the above detailing requirements have been met.

Im still unclear on the logic here though. Rapt and others may be able to confirm if this is the correct approach and shed a little more light on it.

I thought I answered your first question above. Yes, the code says all walls. Not columns. And yes there is a problem with this if engineers try to game the system by calling a wall a column based on some length to width ratio, which is why I tried (apparently unsuccessfully) to introduce some sensible relative stiffness logic into the discussion to explain why.

Mu = 1 controls the detailing requirements. Mu = 2 or 3 or whatever controls the moment and force to be designed for.

So Mu = 1 might tell you that detailing for a column in tension is required, controlling minimum reinforcement (e.g. clause 14.6.7, but not limited to this clause, there may be others) etc, but the compression and tension forces the column actually has to be designed for strength for are from mu = 2.

So mu = 2 might tell you that the column is fully in compression under the (elastic drift) which gives the design ultimate load. But mu = 1 says it is in tension under full plastic drift.
If the reinforcement was detailed based on mu = 2, then the amount of reinforcement supplied might not be sufficient to provide multiple cracks and ductile response under the full earthquake drift. Minimum for compression is .0025 or in a very lightly loaded column .0015.

Mu = 1 however tells you it must be detailed for tension as its extreme drift condition is tension, so the minimum reinforcement is the minimum required for tension to guarantee multiple cracking. This is controlled by 14.6.7 and is dependent on steel and concrete strength and is independent of the magnitude of the applied loads. In the plastic hinge zone, for 80Mpa concrete and 500MPa steel it would be .0125, so 5 times the compression minimum! This then reduces as you move up the building away from the plastic hinge zone.

So your starting point for the amount of reinforcement required is the minimum for tension in this case, not the minimum for compression (actually it is the higher of the 2). The column capacity is then checked for the mu = 2 applied moment and axial.

Hi again, just to keep this thread going...

CL14.4.4.3 Axial load limit for elements with u>1. The description says N*/Ag, where N* is the sum of the seismic weights defined in AS1170.4 (earthquake code). So, is this axial load limit only for G + 0.3Q?

It says increased axial load resulting from vertical ground accelerations, if appropriate, should be included, which would then only be applicable to Parts and Components in Section 8 of AS1170.4 - so maybe non-load bearing precast walls. EDC 1, 2 & 3 state vertical ground accelerations need not be considered.

But should axial load from horizontal load EQ be included in this? I would think so..

The paper by Scott Menegon and John Wilson, 'RC walls in Australia: design and detailing to AS1170.4 and AS3600, Table 3 states the 0.2 axial load limit for N* is calculated using the load combinations for EQ action in AS1170.0, which would be G + Eu + 0.3Q.

14.4.4.3 says Increased axial loads resulting from...system behaviour of the overall structure (eg frame action) should be included in N*.

I think this might be what you're looking for.

The loading is gravity load based on the seismic weight. The frame effects mentioned are frame effects on distribution of gravity load. This was mentioned to make sure that the effects or transfers and sway effects due to non-symmetrical structure and loading were included.

Horizontal earthquake load is NOT included.