Highly-loaded anchor bolts in shear

[JSA2]

I have some very highly loaded anchor bolts ("through bolts") in shear. The service load shear force is 32kips to the worst case bolt of an 8-bolt group. The bolts are "through bolts" through an 8" slab with a thick steel plate at both the top and bottom of the slab. The concrete crushing of the bolts in shear will be restrained by the top plate. (The bolts are shear-loaded from the top of the slab.)

I imagine that ACI 318 Appendix D will be too restrictive and will not provide guidance on this condition. Does anyone know of any resources which would provide contemporary limit states for highly-loaded anchor bolts in shear when the concrete crushing limit state is restrained by the top steel plate?

 

[JSA2]

I think what we are seeing is that this is not a problem that can be addressed with a finite element model because it is not elastic. Even a plastic model of some kind may not be able to adequately predict the ultimate strength and deformation of this connection because we could probably not quantify the effect of the confined and crushed concrete under the plate.

I really believe we need testing, followed up with some kind of anaylysis procedure which is developed based on the empirical testing results.

From a design perspective, I like the idea of added shear lugs to assist the bolts. But again, he first mission here is come up a with a capacity of the connection as designed.

 

[a2mfk]

I don't understand all the heavy lifting you are doing as a peer reviewer, I probably don't understand your role. Usually you'd be just reviewing calcs and drawings and making comments...

 

[JSA2]

a2mfk,

Good question and good point, but my preference is not to demand that the designer redesign if what they are proposing works. I don't know that it doesn't and I suspect that it does, given the right bolt size and plate thickness.

The difficulty with this problema and the reason I brought it to this forum is not to have it analyzed for me but to find the testing and possibly the resulting design approach which I believe is out there. I was pretty much told this by my Hilti rep.

I also believe that the results of this testing would be valuable for all of us in the future, given the limitations of ACE 318 Appendix D on our designs.

 

[a2mfk]

I think I'd do two things if a shear lug is not possible:
1) Design/check it in Hilti's program
2) Assume an edge failure with something like a conservative X degree failure plane off each bolt closest to the edge, and then reinforce the slab using strut-tie methodology.

 

[ishvaaag]

I misquoted the tensile stress in the studs in the model that closed at 40 MPa... it was under 400 MPa. 3900 something kgf/cm2.

Certainly this thing is not elastic etc but give you insight of what happens. The steel constitutive law as elastic-plastic can easily be input to the program,and the analysis be made elastoplastic. On the other hand, if the steel stresses don't attain the elastic recovery point, notional or actual, and whilst the concrete is not showing cracks nor the concrete stresses attain -again- the limit stress for elastic recovery, notional or real, the model can be quite good... if closed; and not to forget that this should be thought more as a simulation that an analysis, once the friction or behaviour of the surfaces kept sliding one over the other is mandated. Useful but never as definitive as test.

The observation of a2mfk on peer review may be correct in what expected by average practice, yet might not be always what expected of the reviewer, or asked of. On my part, I use the questions here posted as much to share anything I may know as to exercise skills and learn.

 

[a2mfk]

So do you have the EOR's calcs? Seems like the first place to start from if you are checking out his design...

This is not your average run of the mill connection, which is probably why you are having difficulties locating research or testing on it. 8" slab is comparable to a tilt wall panel, but thru bolting with sandwich plates is pretty uncommon with tilt. What I am getting at is I'd be surprised if you can find research or testing on something close to this situation. You will be left with bolt manufacturer's tables and programs like you have done, ACI Appendix D, or another approach like I mentioned before. My recent experience with the first two leads me to other approaches once the load gets above a certain threshold...

 

[dik]

FYI

Dik

 

[ishvaaag]

I have resumed in one pdf the 2 elastic models 1. fully bonded and 2. just contact surface (no friction)

 

[ishvaaag]

I make a necessary correction to my comments in p. 12 of the above file, that I upload corrected, since the maximum tensile stresses are not, as could be expected, in the same range in the second model than in the first; they are about ten times higher. I also make some minor cosmetic or spelling corrections.

One professor quoted the adage of that an architect should have more eraser than pencil. I am fortunate in that if I was a native american my name would be "I always correct"

 

[ishvaaag]

at p. 12 also it must say...

"... since compressive strength is 35 MPa"

 

[hokie66]

The AISC recommendation which dik attached suggests using shear lugs rather than depending on bolts because column base plates normally have oversized holes and grout under the plate. If neither of these apply, I see no reason why through bolts wouldn't be appropriate. As to the confinement by the plates on both sides, I would use an analogy to the increase in bearing stress based on ratio of loaded area to total area.

 

[dik]

For large shears, I used to use welded plate washers... now I use shear lugs.

Dik

 

[JSA2]

Thank you all for your contributions.

Ishvaaag, I like your native American name, "I always correct," but when practicing engineering as a business, you're better off with the native American name, "You always correct!"

Thank you for the FEM outputs and your observations and comments on them. I noticed that the plate shear friction modeling would not close on solution and unfortunately, it appears from my limited review of a couple documents yesterday that that is where the ultimate practical solution to this problem lies: shear friction and confinement with the plate.

I believe this problem is essentially a baseplate design problem. Here is what I've found so far. I've only skimmed through them so far and am not sure if they will be useful for this problem, but I wanted to share these with all of you:

-- AISC Steel Design Guide #1 - Base Plate and Anchor Rod Design
-- Shear Transfer in Exposed Columnn Base Plates, Report to AISC March 2009

A third document which does not necessarily address this problem, but has some useful insight on large diameter bolt performance is:

-- Shear Headed Anchors with large Diameter and Deep Embedment in Concrete, by Lee, Park and Suh, from transactions... Toronto August 2007, Paper #H07/2.

This weekend I have billable work to do and will have to continue this study next week.

 

[ishvaaag]

Certainly, as other says, the expert knows. Those that have not the fortune of true expert advice or endowment are left to the path of progressive perfectionment.

Respect the question, I have not tried more but may, I still think the model not closing being precisely a matter of not enough section. On the friction matter, I also have the doubt that it can be as effective as to show the 1.25 inch diameter valid along the lines of this model, whether the materials and analysis are upgraded to nonlinear or not.

I may have also some info on related questions of anchors etc that may also give insight to the behaviour and search for it.

 

[ishvaaag]

See for example the allowable shear capacities for 1 1/4 adhesive threaded bolts included. The capacities listed for 4000 psi concrete suggest the diameter won't be enough for forces between 13 and 16 tonnes even if factored loads.

 

[ishvaaag]

See the evaluation of the loading at the worse bolt by one instantaneous center of rotation scheme, per the letter or former steel EA 95 code (I have dispensed myself for now of establishing the forces in rational way and have just modified one worksheet that I had for the case where the eccentric shear was parallel to the columns of bolts). This workheet should do the same when it is inclined, and if not a correction by rational assumptions would be required. It places the load at the most solicited bolt at 14.8 tonnes or 32.62 kips.

 

[gte447f]

"I believe this problem is essentially a baseplate design problem."

Yes, and that is why several folks have tried to steer you toward the AISC's recommendation that you not rely on the anchor rods, or thru bolts in this case, for shear resistance. It is my understanding that AISC's recomendation is intended to address, or rather avoid, the unknown distribution of shear to the bolts that is a result of oversized holes in the baseplate and high construction tolerances for setting anchor rods in cast-in-place concrete foundations. Some use welded plate washers to essentially "lock in" the shear distribution, but apparently AISC is not a fan of this method. As hokie66 alludes to, if you can be sure of the shear distribution to the thru bolts, then AISC's recommendations don't necessarily apply to this situation.

Crushing of the concrete is an entirely different design criteria which will have to be satisfied, but has nothing to do with the AISC recommendation.

If this is a peer review, why not just request the designer's calcualations to review as a start?

 

[JSA2]

gte,

In answer to your question, the designer's calculations were incorrect regarding both load distribution and capacity of the connection. This got me wondering what capacity could be justified for this connection and why.

Assuming no oversize (standard) holes, which would be easy to do using templates, I believe that the confinement provided by a thick enough plate would result in a capacity greater than that provided for by ACI 318 Appendix D, given that the shear capacities given by Appendix D do not consider this confinement.

My search for this testing only turned up base plate design guidance, of which some of it is useful, but none of it addresses my initial question reqarding the impact of confinement of the lateral capacity of this connection. The testing I did find assume oversize holes with welded plate washers and grout and only conclude that standard holes may be different.

There is, however support for the shear friction concept using a fiction coefficient of 0.45. If shear friction were the mechanism, the concrete crushing, by definition is considered. Normal column baseplates have significant normal force from the column loads, which this connection does not have. I guess I would worry about the deflections in this case.

Regarding the peer reivew project, it is, and has been for about 5 days in the designer's court to correct her loads to the bolts. Once she does this (and thank you Ishvaaag for supporting the 32 kips per bolt. That's what I got too.) we will move on the actual capacity of the connection. All these posts are my journey preparing for this capacity discussion.

Thank you all for your contributions. I will let you all know where this goes from here.

 

[SAIL3]

Your intuition and gut feeling about this connection is well warrented.
Using bolts/anchor bolts for the transfer of large shear loads in concrete is not advisable because of their poor performance.
Same goes for depending on friction.
As others have suggested and is my practice, also, to use shear lugs for any significant shear load.
Here are the questions I would ask:
1....what length of the bolt is actually effective in
transferring this shear..I would say 2 to 3" and based on
that it would result in conc. crushing.
2....what is the profile of the bearing stress on the bolt
around the circumference...would have a peak stress at 90deg
tapering to zero at 0 & 180deg. One of the reasons flat pl
shear lugs are more effective is because of their almost
uniform bearing stress distribution.
2a...can a bolt "travel" thru a chrushed conc zone due to it's
profile?...resulting is larger than normal displacements.
3....how much bending in the bolt?...difficult to quantify.
4....given that there is conc chrushing, is the confinement
by the plates sufficient?...again difficult to quantify.
5....what is the overall affect on the slab that now has a
localized zone of pockets of chrushed conc ...I would have
to check this weakened area?
6....are the loads reversible..if so, that could result
eventually, in unaccepable displacements.
7....are the plates rigid enough to distribute the loads, as
assumed?.

The mounting number of questions/unknowns usually signal that there is got to be a better alternative out there...trying to justify this, as it stands,is like trying to put lipstick on a pig.

 

[JSA2]

SAIL,

Yes, I agree. These are the questions and I believe they could only be answered adequately by a testing program designed to separate out the variables.

This is what started this whole thread. I was hoping to find this testing, but it probably doesn't exist. For my current problem I will have to rely on ACI 318 Appendix D and use the capacity limitations for concrete breakout and steel strength and reinforce the concrete in the vicinity of the anchors.

Thanks for your contribution.