Field Retrofit of Steel Beam Connection

[jdgengineer]

We have a field condition which we'll need to retrofit for a connection between a steel beam and supporting girder. The steel fabricator erected all of the steel before sending us shop drawings for review (we have told them this is not acceptable). As a result, we have an odd condition where the TOS for the beam is ~2 1/2" above the TOS for the girder. Had we reviewed the shop drawings we would have raised the TOS of the steel of the girder and avoided this condition.

The beam is a W8x67 and the girder is a W12x50 (don't ask -- residential construction). To make the connection the steel fabricator notched the W8x beam and welded the web of the beam to the top of the girder. They then installed a notched extended shear tab with (2)-3/4" A325N bolts to the web of the W8x beam. The loads are not extreme (~12k LRFD)

Now that the condition is constructed we are left trying to accept the connection or find a field fix. In my opinion there are currently two load paths:

1) Notched bearing seat
2) Extended shear tab

Due to unknown potential differences in stiffness I want to resolve the connection solely with one of these two load paths (i.e. not combining two). The notched bearing seat almost calcs outs (works for shear but fails in flexure if we assume the reaction point is at the web of the W12x to avoid introducing torsion into the beam). It fails by ~5% so it's pretty close. However, the notch is pretty extreme and almost the entire web is cut out. Even if it calced out I'm not sure how comfortable I would be with it.

Therefore, I'm left trying to get this to work through the extended shear tab. With the notched bearing seat I believe we have lost the rotational ductility of the connection and it should be considered "fixed" now. However, due to the fact it is a single-sided connection introducing torsion into the girder I don't believe the moments would be significant as a result. Therefore, I'm not overly considered about the rotational ductility of the connection as I believe the adjacent girder will provide an essentially "pinned" reaction.

Counting only the extended shear tab RISA Connection says the connection fails with DCR = 3.2. This failure is due to plate flexural failure as well as eccentric bolt failure.

As a fix, I can think of three potential options

1) Weld the backside of the shear tab to the web of the beam. This significantly reduces the eccentricity
of the connection. However, the line weld is not capable of resolving the torsion on it's own. I could then also weld the other 3 sides of the shear tab to provide better torsional resistance. However, in order to avoid plate flexural failure I'd need to assume the vertical force is resolved by the inside weld and the other 3 welds only provide moment resistance. Considering the difference in stiffness between the beam and the shear tab I think it could make sense, but I'm not sure how comfortable I am with this approach. This would be the easiest fix though. Normally, I know welding shear tabs is a bad idea, but due to the inherent flexibility of the supporting girder in torsion it seems to me to be acceptable in this instance.

2) Provide another shear tab on the other side of the web. In order to get the flexural strength of the plate to calc out, this plate would need to be 7/8" thick. The 3/8" plate and 7/8" plate would be able to act in unison and the bolts would be in "double-shear" (sorta). I would design each plate to take the load based on their stiffness (i.e. thickness of plate) and check the bolt reaction based on this stiffness assumption (therefore not truly double shear). The 2nd shear tab would then be PJP welded to the web of the girder / flanges on 3 sides. The bolts would need to be swapped out with longer A490X bolts.

3) Similar to #2 but abandon the bolts and weld web to shear tab.

I'm leaning towards Option #2. Any thoughts?

 

[KootK]

doesn't the torsion have to go through the bolts in an extended configuration?

Absolutely. However:

1) By definition, you have enough rotational restraint at your girder ends that you're able to call the LTB length the physical length.

2) Unless you grind off the flange connection weld, you do not physically have an extended end plate connection. Assuming the moment to be in the bolts when there is a much stiffer moment resisting path available seems questionable.

What do you mean by "oddly, the presence of the full depth stiffener can reduce capacity and disregarding it can be non-conservative"?

A Larry Muir article mentioned this and quoted some research. I assume that the full depth stiffener tends to draw more negative moment to the plate than you might otherwise see.

As you know, the extended plate method is based on a capacity design philosophy. In my opinion, when the connection becomes very messy line this, so does the proper application of the philosophy. Here, you'll struggle to yield your true moment "fuse" because it will involve your top flange seat connection. So establishing your enormous bolt group moment will be difficult. I'd favour an old school KISS approach. Ditch the fuse and just make sure that everything is stronger than it needs to be. Probably much stronger.

If this was your connection, it sounds like you would be ok leaving as is?

Yup. I would sure like to see you get paid for your evaluation time though.



I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[jdgengineer]

Thanks KootK.

Even though all looks good, for some reason I still want to weld the beam to the tab. It just makes me feel better. Would you have any concerns welding the shear tab to the beam web? Likely not necessary, but cost is not an issue (only 2 beams), but the added redundancy makes me feel a little bit better.

 

[KootK]

I get it. An engineer's intuition should trump all else in my opinion. My only concern with the welding would be the rotational ductility but, like you mentioned, that's pretty much shot now no matter how you slice it. Weld away. One thing that I do like about the welding is that I feel that you'll have more convincing load sharing between the two mechanisms that you've got in play.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[HouseBoy]

from KootK : "Would it work if you assumed all of the moment and shear went into the beefy flange of the supported beam? Ignore the stem contribution altogether?"

I'm probably overlooking something basic but....
Load is 12K LRFD
Area of "stem" is (.94" + 1.56") x .57" = 1.425 square inches
Isn't that enough steel of the shear?

jdg - workmanship looks pretty good (as others have mentioned).
5% seems like a minor over-stress IF MUCH OF THE LOAD IS FROM LIVE LOAD (which I fond is often the case in residential design situations.
Just wondering if that would be a valid consideration to allow slight overage in the calculation in this case.

I get into "crazy" residential conditions like this often so, this discussion is very interesting.

 

[jdgengineer]

Hi HouseBoy -- Thanks for the contribution. Yes, area of stem is no problem for shear. The issue is potential bending in the flange due to the eccentric load. However, per calculation I posted a bit ago, the flange does calc for this, so there is a load path there. I'm just not entirely fond of the amount of notched material left. Works on paper, but something seems off to me to notch 75% of the beam and leave essentially only the flange. But with the beefy flange, seems to be acceptable. I've told the contractor about the welding of the web requirement and they didn't bat an eye at it. I feel a bit better about having some alternate viable load paths.

Yes, a lot of the loading will be live load, but it's probably closer to 60/40 split. The floor is relatively heavy due to 1 1/2" gypcrete topping.

 

[HouseBoy]

Thanks jdg.
Along the lines of a previous comment hoping you "get paid for your evaluation" I'd likely consider a more crude approach - If the .57" web can take the shear easily, I'd look at that weld across the top of the W12 and consider that, along with the bolt group as creating a horizontal force couple that would "counter" the eccentricity of the gravity load from the centerline of the W12.

The weld you asked for seems like it would be very effective so, glad the builder was agreeable.

 

[KootK]

I'm just not entirely fond of the amount of notched material left. Works on paper, but something seems off to me to notch 75% of the beam and leave essentially only the flange.

We seem to have arrived at a practical solution so consider the following to be just some esoteric "drift" off into the theoretical.

1) The most nebulous aspect of the flange connection, in my opinion, is resolving how the moment and shear forces in the flange transition back into whole section flange axial forces and web shear. I can think of some checks to be done but it's all very "off reservation" stuff.

2) With the girder stiffener centred below the flange, the flange is really going to cantilever from the girder flange. That will reduce the the beam flange bending forces dramatically.

3) In terms of how much of the section is removed at the connection (75%), I see this as quite analogous to the situation with OWSJ seats. In that instance, we're okay with it but it often represents one of the least robust aspects of the system.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[KootK]

For sport, I FBD'd my way through how I think this would actually work in the wild.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[jdgengineer]

Thanks guys. KootK I think your the one I hope should be getting paid for their time. If only I could invoice the client eng tips answers...

 

[TehMightyEngineer]

Hey, not to speak for everyone else but bill your time as your time; if you want to be fancy, tell them that you did a very expensive consult with a number of world-renowned engineers but as you're such a nice engineer you'll only bill them for your time.

Just remember to return the favor and help someone else out here on eng-tips next chance you get.

Professional Engineer (ME, NH, MA) Structural Engineer (IL)
American Concrete Industries

http://www.americanconcrete.com