b/t limits?

[KarlT]

I am designing a column cap plate to connnect a 7" x 28" PSL wood beam to the top of a 5" x 5" HSS column. The cap plate will be made up of a 7"x 14" bearing plate (to provide the minimum required bearing area for the PSL beam reactions) and an 8" high x 14" long plate on each side of the PSL beam. The pre-fabricated cap plate will be a u-shaped welded steel plate assembly, which will form a saddle for the PSL beams to sit into. I will be providing a couple of 3/4" diameter thru bolts per beam end to resist the uplift forces on the beams. (Note: the typcial cap plate will have two PSL beams meeting at the connection, with the joint at the column CL)

The 8" vertical dimension of the two side plates was determined by the minimum required edge distances for bolting one vertical row perpendicular to the grain plus a bit extra to give a side plate height of 8".

Now to design the cap bearing plate I am taking the beam reaction over the bearing area to get a UDL and designing the cap plate to span 7" between the side plates, based on wl^2/8, to account for the 4.5" part of the cap bearing plate that is overhanging the 5"x5" column. Then I am conservatively treating the two side plates as "t" wide by 8" high rectangluar beams which cantilever 7" beyond the column CL. The Cdn code requires a b/t limit of no more than 200/sqrt(Fy) for legs of angles, etc in compression. This would result in a side plate thickness of about 3/4", which seems like overkill, since the part of the vertical stiffener that is in compression shouldn't really be able to buckle laterally, as the saddle is in a u-shape. Should I need to worry about satisfying the b/t limit in this case for the two side plates?

Any suggestions on how to design this cap plate? My method above is overly conservative, I could also take half the load to each side and calculate the section properties of the L-shape, but the cap plate still needs to span between vetical plates to prevent crushing of the PSL in bearing? Man am I ever going on and on!

 

[ChipB]

ummmmm...side plates are for uplift right? They won't go into compression.

To design the cap pl thickness, I'd ignore the side plate's contribution to the stiffness, and use a typical base plate design procedure using the column's reaction, and the bearing pressure of the wood.

Hope it helps

 

[KarlT]

Thanks ChipB. The side plates will actually experience compressive stresses near the bottom due to flexure from the downwards beam reactions and will also experience compressive stresses near the top when wind is trying to pull the beams upwards. The plates will be fully welded the the cap plate so they will act as a unit.

 

[ChipB]

KarlT,

Treat it as:
Cap pl takes all downward force (compression)
Side plates take all upward force and transfer to cap pl. They (side plates)will be in tension only. These plates would be designed based on the "tearing" of the section.

You've already stated the required dimensions for your cap plate for allowable bearing for downward forces. The uplift forces will be transfered through bolt bearing. This will determine the # of bolts req'd. The uplift force is divided between the two side plates. The weld between the side plates and the cap plate should be sized to transfer this force. You are allowed a 1/3 increase in stress due to transient loads.

I'm telling you all of this because it appears you are new to this. This is typical design for this type of support. You're making it too hard on yourself, but keep on thinking!!

Chip

 

[KarlT]

ChipB

OK, I am thinking about this and I still come up with the same answer....

I see what you are saying and I do agree with you, however following the flow of forces you outlined above still brings me to the conclusion that the side plates will experience some compressive stresses at the top of the plates during uplift loading.

The 5x5 HSS column is not as wide as the 7"x14" u-shaped bracket plate. The bolt locations for the beams will be 4" away from the column CL, which means that the U-Shaped bracket will behave as a flexural member, since the uplift load is at an eccentricity relative to the edges (or centreline depending on how you look at it) of the column.

Even if you sized up the cap plate to be thick enough to take the moment from this eccentricity, the side plates will still experience flexural stresses since the cap plate will deflect upwards at the ends and act compositely with the two side plates.

If the side plates are too thin, is it not then possible to have them buckle between the bolts at the top of the bracket under uplift conditions?

 

[ChipB]

Karl,
They set the LVL in place on top of the cap plate. Cap plate is carrying the load. Then they drill the LVL at the bolt locations for the side plates. Cap plate is still carrying the load. Then they install the bolts. As the cap plate has been carying the load, at this stage, there is no stress on the bolts, therefore, no compressive forces induced into the side plates.

4" is a decent amount of eccentricity if you have an unbalanced loading condition. You could check it for plate rupture, however, I have not seen this as a controlling element in steel framing, and doubt if the uplift forces in wood to steel framing, could induce it. Everytime I've checked, I've had a 3/8" plate to a TS column, connecting the beam to the column. If you're worried about it, use 3/8" thick side plates.

Ideally, you want your resultant force to act through the centerline of the column and bolts. This negates moment being induced into the connection. Where your cap plate is shifted to one side (i.e. exterior wall), you will have to size the cap plate for this additional moment.

If you and a friend are playing tug-of-war with two others, and you all are pulling equivalently on each end, is there any sag in the rope between you two? The answer is no. As a rope can not take compression, and there is no sag, there is still tension in this area. So, no, the side plates will not buckle between bolts under uplift, as this area is still in tension.

If the assembly is to act as a composite section, it gets more detailed in finding the required weld size. You'll get into shear flow (VQ/Ib)and it gets involved. Most people use the simplictic (sp?) approach.

Hope it clears it up for you.

 

[KarlT]

Thanks ChipB for your tenacity and your patience with my stubborness!

Here is my attempt at sketching out what I am trying to say.

Y
|
Pu | Pu
----------------------------|----------------------
^ | ^
|-e--|-e--|
| | |
----------|A---------
| * | * | <--- C
| * | * |
| * | * | ---> T
X------------------|_________|B________|-------------X
| | |
| | |
| | |
Y
Force Pu (uplift at applied at bolt centrelines) causes tensile stress in side plates the Y-Y direction. However, because Pu*(e) causes a moment as well about the column CL(AB), the side plate will have a nominal amount of compressive stresses at the top of the plate in the X-X direction and tensile stresses at the cap plate and bottom of the side plates in the X-X directions.

If you could make up a u-shaped bracket, restrain it from moving in the Y-Y direction at the middle and then pull upwards on the ends you would put the bracket into combined tensile and flexural stresses. This would be similar to welding a cantilever angle rigidly to a plate in say a concrete wall and then pulling up on it. The angle would bend upwards putting the top tip of the angle into compression and the base leg of the angle into tension at the wall.


^ Pu
/| |
/| |
/|------------------------ ||
/| | <--- C ||
/| | ||
/|_______________________| ||________
/|------------------------ ----> T |---------|
/|
/|

I actually took your advice however and sized up the cap plate for the full downwards loads and then I went with 3/8" side plates.

Look forward to your response, if I am totally missing the point please let me know....hopefully I won't need to go back and check all my old designs now!!

 

[KarlT]

A

http://wwww,

stupid formatting!

 

[HgTX]

Try putting a sketch online someplace (if you have such access) and posting a link to it.

Hg

 

[DTGT2002]

I believe you are trying to deal with the bolt hole deforming in the plate as the beam and bolt are pulled upward due to uplift from the wind (or push down due to beam loads). That deformation check should be in your check of the plates adequacy as a tensile elemant.

You are talking about local compressive stress at the bolt holes right?

I would not think this would classify the plate as a compressive element. As you say, there really is no buckling fairule mode for the backside compression of a bolt hole.

Good luck,

and keep talking,

Daniel