AISC Seismic Design Manual 1

[SteelPE]

I have a question regarding the AISC Seismic Design Manual (Which references the 13th edition ASIC Manual).

I am looking at the beam to column connection requirements for an Ordinary Moment Frame. The example 4.4 in the book uses a W18x40 beam connecting to a W12x35 column. During the check for column panel zone shear they use a value for Mu= 77.2 ft-kips. The example states that this is given in example 4.3. I can’t seem to figure out how this 77.2 ft-kip value is derived. Especially since the connection is suppose to be capable of developing 1.1RyMp of the beam. Where does the 77.2 ft-kip moment come from?

I must be missing something because this doesn’t quite make sense to me.

 

[WillisV]

It's the applied moment (from the applied loads) which is given on p4-13 in the initial required beam strength. Panel zones for OMF are not required to be designed for the beam capacity per 11.3.

 

[SteelPE]

So the 77.2 ft-kips is the moment that is derived during the frame analysis?

If the panel zones are not required to be designed for the beam moment then what are they designed for, just the moment that is derived from the analysis (if the above question is correct)?

Sorry for the simple questions.

 

[WillisV]

Yes - moment derived from frame analysis.

 

[SteelPE]

OK, I was trying to design the panel zone shear for 1.1RyMp as show in section 11.2a. I guess this doesn't apply for the panel zone then.

I am trying to help a fabricator figure out whether or not I need doubler plates for some columns. The project was design by another engineer who is requiring the connection be designed for the full capacity of the beam. Absent the loads from the frame analysis (from above), I guess I am suppose to use the full capacity of the beam or column (which ever is less).

This seems a bit harsh considering since I suspect the beam and column sizes were developed for drift vs strength.

 

[JoshPlumSE]

Designing the panel zone for that higher shear (Cpr*Ry*Mp) would be more like a SMF frame than an OMF.

 

[SteelPE]

With the SMF you would have to then comply with the prequalified connections at the back of the manual....which I do not believe we do. For one thing the column sizes do not comply with the local buckling requirements of the SMF.

Like I said, I was just trying to help a fabricator figure out whether or not they needed to estimated doubler plates for the connections. From the information given, you do need to figure that doubler plates are required.... which is going to be a very pricey requirement.

For the structures I design, I usually opt out of AISC 341 as allowed by the code.

 

[nutte]

It sounds like the engineer that designed the building has requirements that go above and beyond the requirements of an OMF. The question (does he really want the panel zone designed for the beam's full capacity, even though it's not required by AISC 341?) needs to be directed at him. Depending on his mood and the relationship with the contractor, he may or may not be receptive to suggestions or reason.

 

[SteelPE]

nutte,

I agree that they requirements goes above and beyond those for OMF. There is nothing wrong with that.... just that it will end up costing a ton of $ to implement if the sections sizes were based upon serviceability and not strength.

At this point, I was only asked to help with figuring out if doubler plates were required or not. Based upon the requirements given on the drawings, they are. I have relayed this information to my client. I have also given them a quick run down with regards to designing the connections for the full capacity of the beam. How they proceed with this information and quote the job is up to them.

If they have a good sales team, knowing this information should not hurt them when trying to win the job.