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Upon reusing an older calculation,
2

Upon reusing an older calculation,

Upon reusing an older calculation,

(OP)
Upon reusing an older calculation, we noticed that the rebar provided for concrete breakout in a Seismic Design Category D region, is insufficient in quantity. The foundation has already been poured. Are there any mitigation techniques that don't involve tearing up the concrete?

Also, when designing anchor bolts for overstrength load combinations, does the ductility requirement still apply?

Thanks!

RE: Upon reusing an older calculation,

Was the calculation incorrect, or is it insufficient per a newer code/specification?

RE: Upon reusing an older calculation,

If you do have to confine the concrete, you could look into fiberglass-reinforced polymer (FRP) external strengthening.

RE: Upon reusing an older calculation,

(OP)
Drawings were incorrect, didn't reflect what was calculated correctly, and wasn't checked for that.

RE: Upon reusing an older calculation,

Can you give us a sketch? Is the rebar being used to intercept the failure cone of anchors?

RE: Upon reusing an older calculation,

(OP)
@canwesteng, yes it is used to intercept the failure cone. It has been developed in only one direction, and As quantity has not been increased for lack of development in the other direction from the failure plane.

RE: Upon reusing an older calculation,

Can you give us a sketch then? Still quite vague, is it a pier, slab, pilaster, what direction is the load.. etc

RE: Upon reusing an older calculation,

pretty robust... how large is your anchor rod and is it high strength steel? Unless there is something special happening, I could have missed it.

Dik

RE: Upon reusing an older calculation,

(OP)
@dik, the bolts are 2" 105ksi steel. Almost 500kip uplift load. So, the (8) #7 bars don't cut it as far as also including development length requirement on either direction of the failure plane. Plus, this is in seismic design category D site, and ductility of bolt isn't met either.

RE: Upon reusing an older calculation,

Quote:

So, the (8) #7 bars don't cut it as far as also including development length requirement on either direction of the failure plane.

Just looking at it.....I'd agree. But with all that tie/confining steel, perhaps there is a chance you could (instead) try to make it work like a embedded RC column perhaps? (Assuming it's tied properly.) I did something similar once myself.

RE: Upon reusing an older calculation,

(OP)
Thanks, all. The way I see this design is:
1. To maintain ductility requirement, divide load by utilization ratio of anchor bolt steel.
2. If concrete fails in breakout, design rebars for the above calculated load.
3. Develop (in this case, Ldh, hooked bar development length) the bars both top and bottom of where the ~33Deg failure plane from bottom of anchor bolt meets the rebar.
4. If depth of concrete restricts development, increase rebar quantity by ratio of reqd Ldh/available Ldh.
It is the last bit of point 4 that hasn't been done.

RE: Upon reusing an older calculation,

Quote (OP)

the bolts are 2" 105ksi steel. Almost 500kip uplift load

Thanks... would have really checked that one...

Dik

RE: Upon reusing an older calculation,

(OP)
@Dik indeed!

Can anyone think of a solution for this that doesn't involve tearing out the concrete? One of my colleagues suggested welding a plate to extend the exist base plate and add post-installed bolts to relieve the stress on the existing bolts and its dependence on concrete. I am not sure how that will work.

RE: Upon reusing an older calculation,

Quote:

One of my colleagues suggested welding a plate to extend the exist base plate and add post-installed bolts to relieve the stress on the existing bolts and its dependence on concrete. I am not sure how that will work.

Sounds like a plan to me. If what is in the concrete won't work.....a extension is likely in order. You'd ultimately be sharing the load between the old and the new.

RE: Upon reusing an older calculation,

How close to 500k are you? The rebar is good for 432k as shown, two legs transfer load to the bottom of the footing. One select one option from ACI 318-14 17.2.3.4.3, if you select d you do not need to comply with a. Is the 500k conservative? Can you lower loads by using the other options in ch. 17? Adding post-installed anchor will not help your condition. If you have access from the side of the footing, you could potential drill new anchor holes and bolt to a plate under the footing.

RE: Upon reusing an older calculation,

Quote (op)

4. If depth of concrete restricts development, increase rebar quantity by ratio of reqd Ldh/available Ldh.

Gawd no. Two reasons:

1) In doing this, you're basically making your intentionally ductile rebar path non-ductile. Bond failure in rebar anchors isn't meaningfully ductile so the A_req / A_prov trick is baloney.

2) Looking at the failure planes involved, the rebar hardly makes any difference to the situation at all.

When working on these things, it's crucial not to bury your head in the sand and blindly follow the APP D stuff. A quick graphical study of the geometry involved should quickly lead one to the conclusion that the rebar provided in this detail is pointless. You essentially need to head down one of two paths:

1) Is the concrete app anchorage okay without the rebar, following ACI 318-14 17.2.3.4.3 option D? If so, you're done and no remediation is required.

2) If #1 doesn't work out, then we need to start contemplating serious remediation efforts that would likely not rely on the U-bars at all, developed or otherwise.



I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Upon reusing an older calculation,

You're only going to increase capacity if you increase the breakout area, so any post installed anchors will need to go beyond the current breakout surface which is very deep. So it doesn't seem that practical.

If you put new bolts in close proximity of the existing you don't increase the breakout surface very much at all, you have to got much further from the existing bolts to make a significant difference to the breakout surface and hence the capacity.

RE: Upon reusing an older calculation,

(OP)
@KootK the concrete is well over 100% utilization for said load. Anchor bolts is just fine. So this is going to not be an easy fix, sounds like!

RE: Upon reusing an older calculation,

Could you use post-installed rebar to reinforce the breakout cone drawn by kootk? You’d probably need a lot of anchors and large embedment to develop the bar on both sides of the failure plane...but if the other option is to remove and replace, it might be economical.

RE: Upon reusing an older calculation,

Quote (SuKaly)

@KootK the concrete is well over 100% utilization for said load. Anchorage is just fine. So this is going to not be an easy fix, sounds like!

Don't be coy, how much over? The only other non-invasive trick I can think of is to base your numbers on 56 day concrete strength and back that up with some cylinder testing. It's slow going making up the difference though as the improvement climbs with SQRT(f'c). If you only need 15% or so, you might get there.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Upon reusing an older calculation,

(OP)
@KootK it is at 350%

RE: Upon reusing an older calculation,

(OP)
@CANPRO Thanks! This was my initial thought, without adding any more bolts/base plate etc.

RE: Upon reusing an older calculation,

Do cores of the in situ structure if you are going to think about actual in situ strength. Cylinders will give you are feel for it, but nothing like testing the actual placed concrete when you are trying to sharpen your pencil.

Can you hydro-demolish a squareish hole to the base and then install a plinth that the bolts lap with and which the plinth reinforcement by virtue of having a bigger footprint than your bolt group essentially increasing the breakout surface. Size plinth and detail plinth size to give you the required strength (drill bars in sideways to the original concrete to achieve a good construction joint. Basically you don't need to demolish the entire thing if it can be helped.

RE: Upon reusing an older calculation,

Ok just saw the 350%, prepare the demolition crew...

RE: Upon reusing an older calculation,

What is this connected to? Where does the 500 kips go - assuming that you have a mechanism to get it out of the bolts?

RE: Upon reusing an older calculation,

(OP)
It is a transmission tower anchor bolt design.

RE: Upon reusing an older calculation,

Have you thought about my "embedded column" idea above? I'm not sure what the ties are....but if they are good....you aren't talking Appendix D anymore. You are talking ripping out by the punching shear equation.

RE: Upon reusing an older calculation,

Yeah, short of having a some kind of Hogwarts, exponentially increasing strength gain curve, you're probably in for a rough ride here.

350% over-stress basally means that you'd need an underside anchor plate the size of a small swimming pool to get this done. The lines blur but, at this point, I'd call this a failure of footing design rather than an anchorage connection failure.

Transmission towers aren't my wheelhouse but is there any way that one could claim energy dissipation via foundation rocking and just try to prove that your anchorage could lift the footing clean off of the ground. It a good enough story for a lot of high-rise buildings apparently.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Upon reusing an older calculation,

(OP)
@KootK yeah, transmission towers are new for me too. Interesting idea on the energy dissipation. I'll do some research. The foundation is actually drilled piers with a 4' pile cap, but unfortunately no piers right under the bolt locations. Seems to me like the failure will be local "ripping off" than whole foundation lifting up.

RE: Upon reusing an older calculation,

(OP)
@WARose, I'll start researching that idea.

RE: Upon reusing an older calculation,

Quote:

@WARose, I'll start researching that idea.

You aren't going to find it anywhere (that I know of). It's a combination of ideas. You get the vertical load from the bolts to the "column" by bearing, then you are in it. After that (assuming you've got the verticals properly tied and the geometry makes sense), it's then a matter of the "column" ripping out by a 2-way shear failure in concrete (depending on where it is relative to the edge).

It's just like a column bearing.....but reverse. I've used it a few times.....and no phone calls yet.

RE: Upon reusing an older calculation,

How far along is the construction for this, just foundation and anchor bolts?

RE: Upon reusing an older calculation,

Quote (WARose)

Have you thought about my "embedded column" idea above? I'm not sure what the ties are....but if they are good....you aren't talking Appendix D anymore. You are talking ripping out by the punching shear equation.

I respectfully object to this strategy:

1) Punching shear is not designed to apply to columns in tension.

2) I see it as irrational to expect this mechanism to somehow be better than a hypothetical rigid anchorage plate plowing through the concrete from the lowest point of anchorage.

3) Mathematically, the app D and punching shear numbers for this appear to be almost identical with the app D numbers coming in just the slightest bit lower.

4) ACI commentary on the App D numbers indicate that even they might be unconservative for anchorages deeper than 25 in.

Quote (WARose)

It's just like a column bearing.....but reverse.

I think that's in error in a very important way. A column in bearing delivers shear via compression struts originating on the far side of the concrete. That's about as good as things can get. A column in tension delivers shear as compression struts originating from the bond stresses transmitted throughout the interior of the concrete. And the bits emanating from the hook elbows. That's considerably less good.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Upon reusing an older calculation,

I don't really buy the anchor bolts bearings into the column analogy. You still need to get that tension to the bars and then the bars can break out like a column. When you bear on top of an actual column, it's compressive force (as a fanned strut) that spreads through the column and applies a pressure over the base. Here, the anchor head bears on concrete, and the fanned strut has no concrete to bear on. You'd have to intercept the strut with rebar and develop that into the column section, and now it turns out it looks a whole lot like intercepting the failure cone from concrete breakout.

Edit - of course koot beat me to it by a few minutes...

Being more productive, the only way out I see other than tearing the concrete up is adding more concrete (ie mass) until your uplift can be resisted by what you've got there.

RE: Upon reusing an older calculation,

Is the seismic load on your transmission tower really that high? I've found seismic to be a small fraction of the loads typically seen from the wind/ice on the conductors.

RE: Upon reusing an older calculation,

What about coring a large diameter hole at the location right through the slab and installing another pile and put the HD bolts directly into the pile. New pile replaces the pad, do it at all support locations to avoid any settlement issues. i.e. do what you do on every other site?

There should have been some shear ties (and suitable longitudinal reinforcement) in the slab thickness to pickup this load and transfer it to the piles you just mentioned.

One other option worth considering might be to use those double headed studs they use for enhancing the punching shear capacity of thinner slabs, but use longer ones and grout into cored holes. Those have the advantage over plain straight bars as having much better/shorter anchorage lengths which is what you need here. Although I suspect by the time you are finished you have turned your slab into swiss cheese with 1001 holes.

All this shagging around and you might find its better to wholesale demo part of the slab and do it right.

One point people have not/semi mentioned yet, is the loadpath to the piles and this interface transferring a similar load in tension depending on the exact arrangement. I suspect this part won't work either given the 350% stress issue for the bolts and if its simply the reverse of what you have here for the piles, albeit with a larger outline of the pile and pile hooked bars to the top of your slab.

I think when you are 350% of the design actions, a bandaid is not going to stick too well, you need to put your hand up and redesign it so it works and probably demolish the entire thing and start again.

RE: Upon reusing an older calculation,

Quote:

(Kootk)

1) Punching shear is not designed to apply to columns in tension.

And this isn't a column in tension. Inside of the punching shear perimeter (above the bearing), it should be in compression.


Quote:

(Kootk)

3) Mathematically, the app D and punching shear numbers for this appear to be almost identical with the app D numbers coming in just the slightest bit lower.

If that is correct, it may not be worth it. But where the "d" (i.e. for the punching shear) starts for that is a bit tricky.

RE: Upon reusing an older calculation,

Quote (WARose)

And this isn't a column in tension. Inside of the punching shear perimeter (above the bearing), it should be in compression.

It's a "tension" column in the sense that it's a force pulling away from the concrete rather than punching into it. In my estimation, that's kinda the thing here.

Quote (WARose)

f that is correct, it may not be worth it. But where the "d" (i.e. for the punching shear) starts for that is a bit tricky.

Well, they're only close to equal if one accepts that punching shear is an appropriate way to look at this. I'm arguing that it's not.

Another thing that bothers me about punching shear here is that you wouldn't have the concentrated flexural steel at the column that UofM research has indicated is so important for punching shear to do its thing. I know the counter argument though: "but we do punching shear on footings all the time and rarely concentrate the bottom steel under the columns!". Yup, hate that too.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Upon reusing an older calculation,

Quote:

(Kootk)

It's a "tension" column in the sense that it's a force pulling away from the concrete rather than punching into it. In my estimation, that's kinda the thing here.

It's creating a confined chunk of concrete that has to be ripped out. It's certainly better than a embedded plate (which is outside the scope of Appendix D as well the last time I looked) that gets used all the time.

Quote:

(Kootk)

Another thing that bothers me about punching shear here is that you wouldn't have the concentrated flexural steel at the column that UofM research has indicated is so important for punching shear to do its thing. I know the counter argument though: "but we do punching shear on footings all the time and rarely concentrate the bottom steel under the columns!". Yup, hate that too.

I would expect a flexural check with this....so I'm with you on that.



RE: Upon reusing an older calculation,

(OP)
@CANPRO I am not sure, I just started this job and looked up old calcs and found these issues, I'm working with the engineer who did the calcs to better understand the issues.
@Atrizzy I have asked the engineer who did the calc to look into the loads as it looks pretty suspect to me. Still waiting on feedback.

RE: Upon reusing an older calculation,

(OP)
Turns out, the load calc was incorrect and I discovered that in the combination, one value was not /2. The uplift is less than 50% of the ~500k and I believe it all works out. THANKS A LOT for all your input. This will surely come in handy for future calcs. Appreciate everyone's time and effort.

RE: Upon reusing an older calculation,

If your concrete was 350% overstressed, I hope the actual load is less than 30% of the 500k!

Your detail you posted of the anchors shows 3 anchors in a single line. Is that a standard anchor pattern for transmission towers? I've been trying to picture what this baseplate looks like and where the anchors are in relation to the tower leg, its been keeping me up at night ahaha.

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