I vote for a standard hook. Make sure you have top and bottom reinforcing in the slab.

The bar needs to be developed. A standard hook may or may not do that, depending on the footing depth, bar size, etc. As per Jed, you need top bars in the heel of the footing.

Neither a hook nor development is necessarily sufficient here in my opinion. At the connection of the stem to the footing, you're creating a moment connection, not just developing bars. Conventionally, the wall bars are made to lap with the toe bottom steel to transfer whatever tension needs to wind up in the toe bottom bars. Often there is no lap as the stem bars are simply bent into the toe bottom bars. There are other ways to make the connection work but, to my knowledge, none is as simple as just developing the stem bars.

You definitely do not want to turn the "hook" towards the heel. It should extend into the toe so that the bursting stresses emanating from the bend will be restrained by a concrete strut.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

The vertical bars need to be vertically developed on both sides of the maximum moment plane (top of the footing). With a standard hook, you need to provide a footing depth of Ldh plus your cover (12 db is not Ldh). I highly recommend reviewing the CRSI design manual for this type of reinforcing detail.

I disagree with KootK, and agree with Teguci. Sketch 1 is no better than sketch 1. Extending the stem bar, after it bends into the toe, beyond the hook extension required by Code does NOTHING to increase the development of the stem bar. The tension in the stem bar cannot go around the corner into the toe, and transfer into the toe bars by lapping.

DaveAtkins

Quote (Teguci)

I highly recommend reviewing the CRSI design manual for this type of reinforcing detail.

So do I. In virtually all of their details, they show the "hook" extending to the toe of the footing as I recommend (see below). If desired, one could lap the horizontal leg of the hook with separate bottom bars.

Quote (DaveAtkins)

Extending the stem bar, after it bends into the toe, beyond the hook extension required by Code does NOTHING to increase the development of the stem bar.

This is a misconception that pervades north american practice. In Britain, for example, it's common to develop bars around corners with the extensions much further than the 12db. One simply needs to check that bearing stresses inside the hook bend are not exceeded. This will likely pan out to be the way that we handle this in north american practice: Link

Quote (DaveAtkins)

The tension in the stem bar cannot go around the corner into the toe, and transfer into the toe bars by lapping.

I disagree and contend that this is exactly what needs to happen.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Huh, I thought CRSI clarified this.

There is nothing wrong with extending the hook (I usually detailed a 6" toe and saw no point to adding another mat of rebar). However, if the hooked bar is not developed a distance of Ldh into the footing, then it is not developed per chapter 12 of ACI 318. I don't know what the British standards are for hooks and radii. It makes sense that larger radii will decrease the splitting stress on the inside of the hook but the OP specifically mentions 12 db and "standard" hooks (ie not non-standard hooks).

As an example, if we were detailing with fy = 60 ksi and f'c = 4 ksi, #6 bars would need more than a 12" deep footing to be developed in.

Quote (Teguci)

but the OP specifically mentions 12 db and "standard" hooks (ie not non-standard hooks).

That really speaks to my point here. This isn't a development problem in my opinion. As such "standard hooks" really have little bearing. The stem dowel that has an elbow in it is not primarily developed rebar. Rather it is rebar transferring moment around a high demand "closing/opening" joint. There's a significant difference there.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

This kind of stuff:

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

KootK's points are well made, and I agree. It is not just a development issue, it is a moment joint.

This may be a silly question, but does monolothic pour versus separate pours have anything to do with the crack location shown in the figures above?

Quote (mike2073)

This may be a silly question, but does monolothic pour versus separate pours have anything to do with the crack location shown in the figures above?

.

Not to my knowledge. Obviously, if there are construction joints, then those joints might deserve direct shear attention of some kind (shear friction, dowels etc).

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

mike20793,
Yes, there are differences. Constructions joints are by definition "cracks". Monolithic construction in the same configuration may or may not crack. Reinforcement is to control cracking, not eliminate it.

In practical terms, rebar doesn't transfer moment, just tension. For the stem, I think we can agree that, to be useful, the reinforcement needs to be developed for its full tensile capacity (a nice yielding failure should the wall ever be overstressed). Therefore, to resist, we need the bar to be fully developed into the footing. ACI 318 tells us that we can get full development with development length, a standard 90 or 180 degree hook or headed/mechanical devices. In the commentary, ACI notes that hooks fail by splitting the concrete on the inside of the bend, followed by slipping of the bar. If we lengthen the tail of the standard hook, does that preclude this type of failure? Not according to the forces on a hooked bar at failure which shows shear transfer on the bar stopping about 4 or 5 bar diameters from the hook with the rest of the tail acting in bearing. We might get a nominal amount of additional capacity, but nothing allowed by code.

As I see it, for a shallow embedment length, the compressive capacity of the concrete will be exceeded before the yielding of the bar. The concrete will split in the plane of the hook, the bar will slip and the wall will fail. The split will be in line with the compressive strut in this "closing" joint as is expected under a strut and tie model. Because the bar has a full development length into the footing, at some point the radius will elongate and halt the compressive failure of the concrete.

For walls taller than 1-story, I usually showed an additional bar on the inside of the hook. Tests showed that this additional bar adds about 10% to hook capacity. Footing concrete is cheap and, for moment capacity, directly trades off with rebar, don't skimp.

Quote (KootK)

Not to my knowledge. Obviously, if there are construction joints, then those joints might deserve direct shear attention of some kind (shear friction, dowels etc).

Thanks. This is what I thought, just didn't know how a construction joint would affect the propagation of the diagonal crack.

Quote (hokie66)

It is not just a development issue, it is a moment joint.

I am not sure what you mean, hokie66. As Teguci stated, per ACI, you cannot transfer tension around a corner in a moment joint. The bars must be separately developed in each member at the moment joint.

DaveAtkins

I feel as though I've done a pretty good job of demonstrating that you can transfer tension around a rebar corner. In countries outside North America, codes direct designers to do this by checking bend stresses and adjusting radii accordingly. In North America, this can be done using the curved bar STM model.

Quote (DA)

the bars must be separately developed in each member at the moment joint

This exactly what you don't want to do. This case would be represented by the left hand sketch in figure 13-27 above (60% calculated capacity).

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Well, I am not sure what Teguci means, so I suppose that makes us even. The Figure 14-1 which KootK provided above shows how I think retaining walls should be reinforced.

The fact that we use the hook extension as the toe flexural reinforcement means that we're pulling on both the horizontal and vertical legs of the bent bar. How would that be statically possible if rebar tension could not be transferred around the bend?

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Some teasers that will encourage forum members to run out and purchase this fine document on curved bar nodes:

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Arthur Nilson is my go to guy for corner and tee joints in concrete structures.

http://www.eng-tips.com/viewthread.cfm?qid=360662 previous discussion on hooks

http://www.slideshare.net/atulkumarengineer/bs8110... access to an old version of BS8110 - equation 50 is on page 83

To get back on track - OP question - Does the reinforcement have to be developed into the toe? No, it just has to be developed into the footing and into the wall. Will a standard hook work? Yes as long as it is embedded Ldh into the footing. Can the hook be turned toward the heel? Yes, but it is bad practice (zero toe walls have been built).

If you'd like to design to British standards, then don't use terms like "Standard Hooks." If you'd like to save a couple of inches on that footing depth and you are itching to publish an engineering dissertation on curved-bar nodes as well as field an extra RFI from the rebar detailer, go ahead with strut and tie modeling.

• http://files.engineering.com/getfile.aspx?folder=07fce983-cf98-4c6b-89af-d7

Forgetting about code rules and logic etc and looking at where tension reinforcement needs to be, we need tension reinforcement down the back face of the wall and in the bottom of the toe and the top of the heel.

I would have top and bottom reinforcement in the footing to provide the last 2 of these. The bars in the back face of the wall then need to develop past the top footing bars and also lap to the bottom footing bars in the toe with sufficient bend diameter to allow force to develop fully past the bend to form the lap with the bottom bars. The horizontal length of the wall bars needs to be in the same layer as the footing bars and below the support footing bars so as to not create a zone where concrete u=is being relied on to take the tension..

Teguci,

If you have a zero toe wall, you have an opening corner joint. Your statement that "a standard hook will work" is incorrect if the joint is to take the full moment. You need to study the details which KootK posted, which list the efficiencies of various types of joint reinforcement. Testing by Nilsson and others has shown that the double hairpins as shown in Figure 13-29 are really the only way of fully developing opening corner joints.

As I have presented methods that seem to be controversial, I thought I might expand a bit upon what I actually do in practice:

1) I don't use eurocode provisions except as an aid to my own understanding.

2) I use the CRSI joint detailing whenever possible (95% of the time) and don't bother with curved bar node STM theory in those instances except as an aid to my own understanding of joint behaviour.

3) When I have cause to stray from the CRSI detail, I use a spreadsheet that a colleague made at my request to investigate the joint using the joint using the curved bar node STM model. It's rudimentary and requires some manual parameter iteration. Maybe twenty minutes per run to set up and fiddle with.

4) When I use strut and tie models -- for retaining walls or anything else -- my goal is never to shave down material quantities. This is partly because, in my experience, STM never does reduce quantities. Rather, it adds to them. The goal with STM, at least for me, is to make my designs safer, not cheaper.

As an experiment, I googled images for "retaining wall reinforcement". About a third of what comes up demonstrates a fundamental lack of concrete joint detailing principles in my opinion. It's a problem.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Figure 13-27 above has enlightened me. I have always followed the CRSI details, except I have typically omitted that diagonal bar. I felt if the bar from the stem wall into the footing is developed in both directions (above and below the construction joint), and the bars in the footing are developed on both sides of the stem wall, then there is no chance of failure.

I think I will start including the diagonal bar in my retaining wall designs.

Out of curiosity--is the problem with Figure 13-27 a strength problem (reduced capacity), or a serviceability problem (joint opens up)?

DaveAtkins

I too will have to consider adding the diagonal bar in 13-27. We have used that detail in watertight structures like one of the figures above, but it is not something we tend to add in a retaining wall application.

http://transportation.mst.edu/media/research/trans... - further reading on the history of ACI standard hooks as well as new testing regarding their effectiveness at different angles (ex towards the toe or towards the heel)

So again, back to the OP, yes a standard hook developed into the footing will provide full tension development of the stem wall reinforcement. This standard hook at the development length has been tested to its yielding capacity, has been codified by ACI, and has withstood years of real world scrutiny. Pardon our North American misconception.

Please mind that I am not talking about compression failures in the footing or proper footing reinforcement. That was not your question, but I do argue that increasing the depth of the footing for Ldh will have a positive impact on those issues. Further thoughts - Do we provide a full vertical lap splice for the hooked end of haunch/corbel reinforcement? How about for moment connections at beam to columns? Edge beams, pile to pile caps, etc...

Yes indeed, interesting and illuminating thread.

Koot, those images you found are awful, the stuff nightmares are made of, frankly. I wish I could say I haven't seen drawings with the likes of such, but I have.

Regarding Fig. 13 (a), the plan view of the walls, I can't imagine anyone designing corner reinforcing that way, save for my previous sentence. I can easily imagine them getting built any of those - or other - ways, though. Which leads to my next question...

Regarding the diagonal bar in Fig. 13 (b) has anyone found them to cause difficulties with placement or consolidation? If so, I can see how on a large project the issue could be addressed but I can also see how on a small project they might get intentionally omitted at the site.

Quote (Dave Atkins)

Out of curiosity--is the problem with Figure 13-27 a strength problem (reduced capacity), or a serviceability problem (joint opens up)?

It's definitely a strength problem as the efficiency percentages quoted indicate [Moment capacity achieved in testing / Moment capacity calculated by conventional procedures * 100]. The crack widths quoted are, I believe, the values estimated to accompany a serviceability load level (55% Mn for example as indicated). At serviceability crack widths of 1/4", it's pretty tempting to call that a failure as well from the perspective of corrosion and water tightness (where required).

Sadly, I do not yet have Nilsson source document (Link). Might have to rectify that. If you google the document title however, you'll come up with plenty of derivative papers:

Link
Link
Link

I've known engineers who compensated for the efficiency percentages by simply adding that much more rebar. I'm skeptical of that approach, however, as I'm not convinced that it's a linear thing.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (Dave)

Regarding the diagonal bar in Fig. 13 (b) has anyone found them to cause difficulties with placement or consolidation

I'm assuming that we're talking about the wall corner detail below. I've only attempted it in large scale industrial applications where there was plenty of room and there were no complaints. I would wouldn't be pitching it in your standard 8" stem wall or anything like that.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Actually I was thinking to their use at the base of the stem of a cantilever retaining wall. With vertical concrete placements sometimes diagonal bars can case problems. I'm wondering if anyone here has dealt with that in this circumstance. I agree that (per the studies) they sound like they're what's required to control cracking at that joint. However, if it proves to be too difficult to build it will get left out more often than not.

Quote (teguci)

Do we provide a full vertical lap splice for the hooked end of haunch/corbel reinforcement?

I do in fact. Obviously, though, I realize that there are older, empirical methods (ACI shear friction etc) that do not require this. Strictly speaking, if you approach it from a strut and tie perspective, it's tough to avoid the need as you wind up with another of those situations where you need to transfer rebar tension force around a bar bend. The sketches below are taken from the curved bar node article that I posted above and MaGregor's text.

Quote (teguci)

How about for moment connections at beam to columns?

Absolutely at roof level edge columns. We don't typically bother at other conditions where there is beneficial clamping action available from columns both above and below. That said, I'm sure that these connections deserve more attention than they typically receive. And in seismic country, they get it. The sketch below is taken from MacGregor and illustrates the demand.

Quote (teguci)

Edge beams, pile to pile caps, etc...

Oftentimes yes. I'd need more information on the specifics to respond in any meaningful way however.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (Dave)

Actually I was thinking to their use at the base of the stem of a cantilever retaining wall.

I never get push back from commercial / institutional builders and quality seems to be fine. Residential guys tend to complain about it so, since these are usually pretty small walls, I just over design the connection and don't bother with the diagonals. Once nice feature of the STM method is that it demonstrates sufficient strength, if not crack control, without the use of the diagonal bars.

To be honest, the biggest problem with the diagonal bars is when my competitors don't include them and make me look bad. Damn competitors.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

>>>I never get push back from commercial / institutional builders and quality seems to be fine.<<<

Hmm, ok, no reason not to call for them then, I suppose. Good to know. Thanks.

http://www.eng-tips.com/viewthread.cfm?qid=306994

Wow, broken record with some of the same characters. I see Hokie can change his mind, but I'm still stuck.

Teguci,

I do change my mind, but haven't in this instance. In the thread you referenced, see my comments about Nilsson's work with opening corner joints.

This is the same concept as a ground floor slab supported by a basement wall below - to count on continuity/slab neg moment you don't just have std hooks each way. We show very outside face wall bars that bend down and extend a full tension lap into the slab. Same deal.

Hokie,
Do you know if CRSI corrected the Ldh requirement?

Thanks for all your input, it has been very informative. Just want to clarify my questions and checked if the revised sketches (attached) are acceptable or required per ACI 318-11. The Embedment Depth into the footing and standard hook lenghts are provided per ACI 318.

1). Figure 3: This figure is similar to Figure 14-1 in the CRSI Design Handbook. Is extending the flexural reinforcement from the wall all the way to the end of the toe, required?

2) Figure 4: Is providing a standard hook, which is not spliced to the bottom reinforcement acceptable? The flexural reinforcement has been designed to resist the lateral loads and has the appropriate embedment into the footing.

3) Figure 5: Is providing a standard hook, which is spliced (with appropriate splice length) to the bottom reinforcement acceptable? The flexural reinforcement has been designed to resist the lateral loads and has the appropriate embedment into the footing.

Thanks, really appreciate your feedback and help.

• http://files.engineering.com/getfile.aspx?folder=1056cbf8-a4bf-412b-9e5e-51

sketch 3 except move the bottom bar to the top of the footing (2" clr) and verify that it is fully developed Ld by the time it gets to the wall reinforcing (hook at the front of the toe otherwise). There is no reason to have a second layer of bottom reinforcement if you extend the hook length to the end of the toe.

@CWEngineer: I'm glad that you're back and that all this debate hasn't been for naught! My opinions are indicated below. Obviously, a consensus is unlikely.

Quote (CWEngineer)

1)Figure 3: This figure is similar to Figure 14-1 in the CRSI Design Handbook. Is extending the flexural reinforcement from the wall all the way to the end of the toe, required?

In general yes. However, if you plan to turn the hook towards the toe and lap it with your footing bottom steel, then I would say no.

Quote (CWEngineer)

2) Figure 4: Is providing a standard hook, which is not spliced to the bottom reinforcement acceptable? The flexural reinforcement has been designed to resist the lateral loads and has the appropriate embedment into the footing.

In general no. Not unless the standard hooked bar is somehow also able to satisfy a) the STM requirements discussed above or b) lap with the bottom bars for the tension needing to be transferred.

Quote (CWEngineer)

3) Figure 5: Is providing a standard hook, which is spliced (with appropriate splice length) to the bottom reinforcement acceptable? The flexural reinforcement has been designed to resist the lateral loads and has the appropriate embedment into the footing.

Yes.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Teguci, I don't know. The Nilsson work was only published in 1973, so perhaps they haven't had time. And I don't think TXStructural works for CRSI anymore.

KootK, about your answer to CWE's Figure 3...I think you mixed up "toe" and "heel". The way he shows it, toward the toe, is correct.

CWE, does any standard hook provide a tension splice length? I thought not.

I'm not switching sides or anything but I am going to make a point that runs contrary to my own opinions in this matter. Per the sketch below, consider:

1) A retaining wall rotated 90 degrees to look like a beam-column joint.

2) Don't sweat the construction joints being the wrong places. All resolvable via direct shear checks.

3) This wouldn't work for "toe-less" walls or short toe walls.

If this were a beam-column joint, I believe that:

1) Most of us would find it satisfactory by inspection.

2) Other than in high seismic zones, I don't believe that checks beyond Ldh would be requied.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (hokie)

KootK, about your answer to CWE's Figure 3...I think you mixed up "toe" and "heel". The way he shows it, toward the toe, is correct

It's not a misprint but I could have said it better. I meant that, should a bar developed to the toe be replaced with another kind of hooked bar, that hook would also need to be turned towards the toe.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

We would accept the beam-column joint because we only want a nominal bending capacity at that location. It is not typically a cantilever...but if it is a cantilever, it should be treated similarly to the retaining wall. I wouldn't consider that clamping enters into it, but then you know my attitude to shear-friction.

There is good information in this CRSI document that recommends hooking the reinforcing inward. See Page 14.
http://www.conveyinc.com/things/crsidesigner.pdf

There is also good information on build-ability for special shear walls, beam column joints.

Good reference, slickdeals. But just turning the bars in only develops "nominal moment strength", according to the caption beside the picture. It is good to see that they referenced the "Research in Sweden many years ago", which I think is the work of Nilsson et.al. which we have been discussing.

The explanations given by @Koutk are convincing. However if the footing is too thick compared to the stem, in a way to ensure a full development length, does the moment connection still matter or not?

Quote (Lak20016)

However if the footing is too thick compared to the stem, in a way to ensure a full development length, does the moment connection still matter or not?

It still needs to be a moment joint but it can take a different form that, while not particularly efficient, may not require the vertical reinforcing to round the corner of the joint. See the strut and tie representation below.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (koutk)

See the strut and tie representation below.

If the potential shear issue is well understood, I am still confused about the second potential issue( on the left side of the draft) Could you explane to me more, please?

In the case that we have a shear wall connected to a thick mat, should the vertical reiorcment round the corner or not?

Quote (lak20016)

If the potential shear issue is well understood

I don't think that it's well understood at all. Consider:

1) Just look at the diversity of opinion here.

2) I made up my own STM for this and have yet to see another in print other than similar stuff for post-installed rebar (attached).

Quote (lak20016)

I am still confused about the second potential issue( on the left side of the draft) Could you explane to me more, please?

I'll try:

1) On the left, your effective flexural depth is reduced to correspond to the level of embedment of the wall stem bars. Normal design practice would use the full "d" of the footing.

2) On the left, you have a concrete anchorage situation with regard to the wall stem bar embedment. It kind of depends on how one chooses to look at it but I view it as a sort of shear failure left of the stem that would involve a depth less than the full depth of the footing.

3) At the leftmost node on the bottom, there's a bursting tendency that exacerbates the demand on the "tie" at that location. In most retaining wall designs, the "ties" will be non-existent as rebar. Rather, the ties will represent footing diagonal tension shear (Vc). Thus, ultimately, we hav eanother source of shear issues.

I've attached a few pages from a clever document from Hilti Europe for post installed rebar situations. It covers a lot of the same concepts. I'm hoping that you will find it helpful.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

In addition to the development requirement, ld, ACI 318-11 12.10.3 requires developing the reinforcement at least a distance of "d" past where it is needed. In this case, d is for the wall and effectively requires the wall reinforcement to extend at least about the width of the wall into the footing. Further, if the shear is close to the concrete shear capacity in the footing or slab, then the flexural reinforcement is not allowed to be cut-off in the tension zone (12.10.5).

As for the strut tie models, I continue to have problems at the tension tie terminations as well as dealing with the unreinforced, concrete shear/tension ties. With the understanding that a bar being developed transfers its tension into the concrete via shear then something like this can be proposed (and also adds credence to the shear being the limiting factor). Without the shear tension ties, many of the strut tie models above and published in books do not work because the tension terminations are shown way to close to the face of an element to allow for the correct tension development even with a hook.



Update - made shear ties orthogonal and shear struts diagonal.
Below is my attempt to show why tension around the bend doesn't exist for the open corner condition. The tension in the reinforcement is effectively 0 when it gets to the hook/bend.

Quote (lak20016)

in the case that we have a shear wall connected to a thick mat, should the vertical reiorcment round the corner or not?

I forgot to respond to this bit of your question. For out of plane wall moment, I would say yes, the rebar generally should round the corner. For in plane "wall shear" moment, the problem is more one of rebar tension anchorage rather than transferring moment around the joint. In that case, rounding the corner with more bar extension than a standard hook would generally be unnecessary. All that said, there's usually more than one way to accomplish any particular goal in structural engineering. The sketch below, taken from MacGregor's text is basically the situation that you describe for out of plane wall moments. Just rotate the diagrams 180 degrees.

Quote (teguci)

In addition to the development requirement, ld, ACI 318-11 12.10.3 requires developing the reinforcement at least a distance of "d" past where it is needed. In this case, d is for the wall and effectively requires the wall reinforcement to extend at least about the width of the wall into the footing. Further, if the shear is close to the concrete shear capacity in the footing or slab, then the flexural reinforcement is not allowed to be cut-off in the tension zone (12.10.5)

In my opinion, these sectional method design requirements could be waived for a member designed via strut and tie and satisfying the various anchorage requirements associated with that method. In all likelihood, of course, the STM anchorage requirements would be more severe.

Quote (teguci)

Below is my attempt to show why tension around the bend doesn't exist for the open corner condition. The tension in the reinforcement is effectively 0 when it gets to the hook/bend.

1) Just to clarify, this argument would only apply to the case where there is no footing toe extension beyond the wall. With the toe extension, I believe that it's a different story and there is definitely tension force in the bar as it rounds the corner.

2) For the joint that you drew, and presented macroscopically, I would agree that there is no force in the bars around the bend. However, research (some quoted above) has clearly demonstrated that having hooks, hoops and such drastically improves the performance of such joints. Logically, that could not be the case if there was no "tension around the bend". The tension around the bend is a static necessity for rebar details like this to perform.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (http://www.eng-tips.com/userinfo.cfm?member=KootK)

2) On the left, you have a concrete anchorage situation with regard to the wall stem bar embedment. It kind of depends on how one chooses to look at it but I view it as a sort of shear failure left of the stem that would involve a depth less than the full depth of the footing.

I've found this detail closed to current post?

However, I've been facing a big issue to find out a scheme of STM of a wide shear wall embedded in cap pile so ; need I build STM to go through this issue?

if apply this detail requirement , is the equilibrium satisfied or not?

• http://files.engineering.com/getfile.aspx?folder=cf453e03-a42a-4fbf-977c-b6

I would recommend starting a new thread of your own and including a sketch of your situation. I'm having a difficult time visualizing your condition.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (teguci)

I highly recommend reviewing the CRSI design manual for this type of reinforcing detail.

I would like to purchase this manual. When I look at the CRSI website, I don't see it. Can someone please show me where to buy the most up-to-date copy of this. Thanks!

EIT

The retaining wall bit of the CRSI manual lives here now: Link

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Thanks KootK, got it! It's a shame that it's divided up into smaller sections. I was looking forward to additional "surprises" in the manual. I suppose I could purchase a superseded edition.

EIT

Quote (Deadblow)

It's a shame that it's divided up into smaller sections.

Oh... don't get me started. I own every version of the CRSI manual ever printed. The new format messes with the continuity of my collection terribly.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

KootK, where did you find the figures 13-27, 13-28, and 13-29 that you posted on January 14th? Thanks

EIT

@Deadblow: Link

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Thank you!

EIT

I'm glad this resurfaced. I've always followed the CRSI detailing but I just don't get how you can have a joint (stem to footing) that has bars developed on both sides of the joint and should satisfy your basic design requirements (all seems perfectly acceptable) but yet it only achieves 70% of the design capacity. I must be missing something.

EIT
www.HowToEngineer.com

Quote (RFreund)

I've always followed the CRSI detailing but I just don't get how you can have a joint (stem to footing) that has bars developed on both sides of the joint and should satisfy your basic design requirements (all seems perfectly acceptable) but yet it only achieves 70% of the design capacity. I must be missing something.

Are you really looking for additional clarification here are you just venting a little concrete joint design frustration? If you need more than the miles of posting above to convince you, I can give it a whirl in more general terms. I'd need to take the kid gloves off though.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

KootK is the "Muhammad Ali of concrete rebar detailing."

DaveAtkins

Rope-a-dope. Come 'git some.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Great thread and contributions by KootK. Certainly some great info here to take on board.

Missed some of these responses...

I think there is plenty above that gets the point across (great job by the way) but I guess one question I have:

When you presented the case of the wall turned 90 degrees with the cantilevered beam case. If you develop the bars does your beam have the calculated capacity (per code)? Or would it be 70% of calculated strength?

EIT
www.HowToEngineer.com

At these closing joints, does a large radius bend (which limits bearing stresses) give higher capacity in practice than a normal bend? Any good test results out there?

Also, where a normal cog/bend is used we generally put a crossbar in the corner of the cog, whereas most of the images above don't bother. Any thoughts on cross-bar vs no cross-bar?

Tomfh: The presence of the bar has either no effect, or negligible effect, from what I have seen in papers on the matter. It is a convenient bar to have for tying steel, nothing more. Somewhere I have a copy of an ACI report that talk about it... *looks suspiciously at structural library*

EDIT: I was wrong; See below. Don't let this be your take-away.

I am too busy for a thorough search (hopefully someone else will have something difinitive), however have a look at this:

Well am I ever glad I couldn't put my hand on that old ACI report. Apparently I was dead wrong:

The latest research I can find is out of South-East Asia and Australia. "Standardisation of hooks in AS 3600" by A. Wheeler, R. Bridge and W. Marsden has a specific section on this exact matter. I quote: "The effect of the transverse bar at the internal diameter of the bar is significant with higher failure load observed in the 10mm bars. For the 12mm and 16mm bars, only the short hooks with no transverse bar demonstrated a pull out failure. In the corresponding tests where the bar was present the bars fractured in the 12mm and 16mm tests."

From my read of their paper, it seems that the concrete crushing limit state governs for small diameter bars (with their correspondingly small hook diameter). As a result, the presence of the transverse bar within the hook spreads the forces out, lowers the stresses, and potentially prevents the initiation of the crushing which otherwise would unzip the bond and allow the hook to pull out.

I suspect this has never become an issue in practice because these hooks often occur at supports, where secondary forces (bearing stresses, etc) act to confine the hook and increase the capacity. That, and I think most practitioners have been detailing the transverse bar within the hook from either their training or gut feel. I certainly do, even after I thought I had learnt it was useless.

Take way: Small bars with tight diameters really need this transverse bar within the inside diameter included. We need to watch out for this detailing in corbels, trusses, and other small but highly stresses bar locations. You really do learn something every day; This makes me very glad to be back here poking around once again...

Quote (RFreund)

When you presented the case of the wall turned 90 degrees with the cantilevered beam case. If you develop the bars does your beam have the calculated capacity (per code)? Or would it be 70% of calculated strength?

Frankly, I don't know. I raised that as an interesting counter example to my own arguments in this thread. If you flip through this ACI doc, it's hard not to come to the same conclusion as we have for the retaining wall joints: bar development is usually necessary but not always sufficient to constitute an adequate joint. In practice, this is what I have been doing:

1) Moment frame joint = detailed design.
2) Cantilevered beam with no back-span = detailed design.
3) Roof beam with no columns above = detailed design.
4) Most other cases = provide Ld/Ldh and hope for the best.

I've always thought it interesting that, if you examine a typical joint from an STM perspective, providing only Ld/Ldh into the columns would almost always result in problematically steep strut geometry. As I mentioned previously, I figure that the beneficial clamping that one gets from having a column above improves that situation.

Along the same lines, I see incompatibility with how we design our columns. We design them assuming the smallest possible sliver for the compression block because that yields the most favorable results. Logically then, again from an STM perspective, one would think that beam bars would need to be developed over the column compression zone rather than across the entire width of the column. But that is not what is done.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (tomfh)

At these closing joints, does a large radius bend (which limits bearing stresses) give higher capacity in practice than a normal bend? Any good test results out there?

A gentler radius does improve matters. This harkens back to two items that I mentioned above:

1) European provisions where bend radius is sometimes "designed" and'
2) The curved bar node STM method where, again, radius is a design parameter.

Quote (tomfh)

Also, where a normal cog/bend is used we generally put a crossbar in the corner of the cog, whereas most of the images above don't bother. Any thoughts on cross-bar vs no cross-bar

My thoughts:

1) Almost everything above is concerned with opening joints. Closing joints are much less problematic.

2) I believe that the cross bars do in deed help. That said, large flexural bars would need correspondingly sized cross bars to have the same proportional benefit. A 10M cross bar won't do much for a 30M flexural bar etc.

3) In my opinion, the cross bars essentially improve bar development/anchorage. As discussed above, development/anchorage tends to be just one of several possible failure modes that need to be addressed in the complete design of a concrete joint.

4) Particularly with larger flexural bars, I worry about the accuracy of placement with the cross bars. For obvious reasons, they tend to wind up at the point of curve tangency with the horizontal leg rather than at the curve mid-point.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

@KootK - thanks for the response! When you say detailed design do you use STM?

EIT
www.HowToEngineer.com

I use the ancillary ACI docs whenever applicable. STM is neat but it takes too damn long unless I have a spreadsheet to lean on which, usually, I don't.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (kootk)

A gentler radius does improve matters. This harkens back to two items that I mentioned above:

1) European provisions where bend radius is sometimes "designed" and'
2) The curved bar node STM method where, again, radius is a design parameter.

Do you know of any practical tests? Intuitively I'm not totally convinced a gentle curve would be that much stronger. Yes you get lower bearing stress, however you need to start the bar bending much closer to the surface, where confinement wont be as good. I'd love to see some destructive tests of the two cases.

Quote (KootK)

1) Almost everything above is concerned with opening joints. Closing joints are much less problematic.

I was thinking of the closing joint case, e.g. your post CURVED BAR NODES 14 Jan 16 21:42, with the strut tie model.

Quote (KootK)

In my opinion, the cross bars essentially improve bar development/anchorage. As discussed above, development/anchorage tends to be just one of several possible failure modes that need to be addressed in the complete design of a concrete joint.

This is how I imagine it. I think the bar would rather get its fingertips around a bar, but again I haven't seen any real test results.

Quote (Tom)

Do you know of any practical tests?

I know of this one. It requires one to read Swedish however.



I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

https://www.concrete.org/publications/internationa...

Quote (CELinOTTOWA)

Standardisation of hooks in AS 3600

Thanks for the link. This includes some useful info alright. My take aways from it:

-Transverse bar does improve pull out capacity, especially for small round bars.
-One you are into real bars (say 12mm deformed), a standard hook with its corresponding small radius bend is sufficient to break a bar, even though we only use 50% of that in practice. Which raises the question - why bother with these big radius bends designed to limit bearing stresses if we know a tight bend will break the bar? Why not maximise the length of straight bar into the concrete?

Kook has made excellent arguments.
Note that 2006 and 2010 Caltrans standard retaining wall details show larger radii for the stem to toe reinforcing hook. If you can look beyond the politics and disfunction of Caltrans, realize that they still have more than a few very good senior engineers and good standard details, with a lot of full scale testing. They obviously changed the ACI std hook detail with intention.

ATSE,

Do you know why they changed to large radius? What do their results show? Stronger joint? Better ductility? Etc

I'm interested to know why gentle bends are so beneficial when tight bends alone happily break bars.

Tom,
Not sure about the 9" radius - seems very large. But 3 diameters (ACI 318 Table 7.2) seems fairly tight.
Kook and others above can articulate the "why" better. No doubt that concrete crushing stresses are lower at the inside of larger radius bend.

Quote (ATSE)

Note that 2006 and 2010 Caltrans standard retaining wall details show larger radii for the stem to toe reinforcing hook.

Thanks for sharing the Caltrans detail. It's really the first time that I've seen evidence of anyone in North America considering bar diameter in practice.

Quote (Tomfh)

I'm interested to know why gentle bends are so beneficial when tight bends alone happily break bars.

Perhaps it's not a big deal. Certainly, I'm no expert on whatever body of related testing exists out there. That said, I see the following problems when it comes to applying the Wheeler results to the retaining wall condition.

1) Wheeler looked a 135 hooks rather than the 90 hooks that we use in retaining walls. That, presumably, was because Wheeler seems to me most interested in the rapid development of beam stirrups.

2) Because Wheeler's primary interest was beam stirrups, his testing regimen stopped at #6 bars. Concrete crushing inside bends is known to be more critical with larger diameter bars as one might encounter in larger retaining structures.

3) Wheeler did not test hook anchorage in tension zones. The area where stem bars enter footings is a flexural tension zone in many applications.

Probably the most important difference is simply this:

1) In a simple anchorage situation, you've got 1 x As X fy contributing to concrete bearing stresses.
2) In the retaining wall situation, you've got 2 X As X fy contributing to concrete bearing stresses because you're yanking on both ends of the bar.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

"In the retaining wall situation, you've got 2 X As X fy contributing to concrete bearing stresses because you're yanking on both ends of the bar."

How so? I do not see a difference... A developed bar cannot have double the stress in the bar. Are you sure you've thought this one through?

Quote (CELinO)

How so? I do not see a difference... A developed bar cannot have double the stress in the bar. Are you sure you've thought this one through?

Pretty sure. The point that I've been somehow failing to make since January is that the joint mechanics are not about rebar being developed. Rather, they are about rebar transferring tensile force around a corner. In the general case where entry and exit bar tension is unequal, some force transfer does occur via bond stress. That's just a small part of the greater story however.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

I think you are doubling your forces for no reason. If we are talking about whatever force the bar carries when entering the joint, I agree, but the way you show it is more analogous to a detailed pulley problem (where you include friction and other internal forces).

Whetherwe are talking about a developed bar or a bar who's total forces are going around a corner (at which point As isn't really representative), you always have equal and opposite reactions.

Yes, I understand the internal joint does take out some load in compression and/or bar development. I still think you have the wrong end of the bar on this subtle detail, but must commend you for your masterful treatment of the subject otherwise (above).

Going to give this one further thought...

I think this is what's happening:



And I think the fact that we are all so used to showing the stress in a bar as AsFy is actually standing in the way of a productive conversation. Unless I'm mistaken, I don't think you're saying that both ends of the bar carry AsFy, but that's what you showed. What you're getting at is that you see the corner "take out" the load. I think you're both right and wrong...

The corner is subject to the stress which the bars carry. The bars still manage to get around the corner, and they carry the load that they carry due to the applied actions. The applied actions can be thought of as coming from the Resistance, or (as I have shown it) from the Input Forces. While I agree wholeheartedly and enthousiastically that the stress in the concrete is roughly 2xFxCosTheta, I do not see that as being the forces carried by the bar.

Do you see the distinction? I appreciate that we are splitting hairs, but I think it is important. Also, you like a debate and I love to refine my understanding of Structural Engineering.... Thoughts?

I agree with CELinOttawa that the tension load is resolved by equal and opposite reactions. There isn't a doubling of the load.

Quote (KootK)

1) Wheeler looked a 135 hooks rather than the 90 hooks that we use in retaining walls. That, presumably, was because Wheeler seems to me most interested in the rapid development of beam stirrups.

I agree this is a factor.

Quote (KootK)

2) Because Wheeler's primary interest was beam stirrups, his testing regimen stopped at #6 bars. Concrete crushing inside bends is known to be more critical with larger diameter bars as one might encounter in larger retaining structures.

Is it? According to the AS3600 tests the bigger bars worked better. Look at the graph. The bigger bars killed it - even with a short extension length. It was the small soft bars that needed the cross bar.

Quote (KootK)

3) Wheeler did not test hook anchorage in tension zones. The area where stem bars enter footings is a flexural tension zone in many applications.

I agree this is important.

Actually I now think Kootk is entirely wrong about the development into the footing. I'll do a sketch and send it through later today... There is no doubling of the force at all.

Quote (CELinOttawa)

Actually I now think Kootk is entirely wrong .

Now I'm really confused. Didn't you think that already? :)

Not entirely! Lol... I thought there might be an influence of the bar continuing past the corner, but now I do not...

Here:

Quote (CELinO)

...but must commend you for your masterful treatment of the subject otherwise (above)

Thanks. Not sure how much of this still holds but, regardless, it was kind of you to say.

Quote (CELinO)

I agree, but the way you show it is more analogous to a detailed pulley problem (where you include friction and other internal forces).

Yes, the joint is in fact very much like that.

Quote (CELinO)

Unless I'm mistaken, I don't think you're saying that both ends of the bar carry AsFy, but that's what you showed.

That is correct. The tension on the vertical leg could potentially take on any value from zero to As x fy. Same goes for the tension on the horizontal leg. I would argue that, in many conventionally proportioned retaining wall situations, both the vertical and horizontal tension forces would approach As x fy (economical proportioning).

If you examine my sketch in greater detail, you'll see that I included a starred note at the top indicating that I'd shown a simplified condition and acknowledging that As x fy represented the upper limit of applied bar tension (equal utilization. I also said as much in the verbiage accompanying that post:

Quote (KootK)

In the general case where entry and exit bar tension is unequal, some force transfer does occur via bond stress.

Let's not get hung up on the 2X business as that is not the salient point here. The salient points here, in my estimation, are these:

1) The concrete bearing stress that would accompany a standard, development only situation, would reflect a zero value for horizontal leg rebar tension.

2) Your typical retaining wall closing joint will have a non-zero value for horizontal leg rebar tension.

3) The non-zero value fro horizontal leg rebar tension in #2 will add concrete bearing stress in addition to that associated with #1.

4) #3 implies that, just because a standard hook is good enough to break the bars in a pure development situation, standard hooks may not fully address concrete bearing issues in a closing joint situation.

Quote (CELinO)

Do you see the distinction?

I'm afraid that I do not. As I see it, flexure in the footing demands horizontal bar tension at the joint just as flexure in the stem demands vertical bar tension at the joint.

Quote (CELinO)

Also, you like a debate and I love to refine my understanding of Structural Engineering....

I know it. You've been missed.

Quote (Tom)

Is it? According to the AS3600 tests the bigger bars worked better. Look at the graph.

Not sure to be honest. I don't have any readily available research to support my claim. Nor can I remember where I picked up my assumption. Before proceeding further, I should mention that I only have access to a partial google doc on this: Link. I don't have the original proceedings document nor Wheeler's original research paper. If you're working from better information, do let me know.

The data that I have access to is shown below. To me, it appears that the larger bars did not outperform the smaller bars. Rather, all of the deformed bars tested (#4, #5, #6) reached the same ultimate stress. That being the breaking stress of course. Moreover, because the failure mode was generally bar tensile fracture, the tests don't really say anything about differences in concrete bearing stress. One would either need to measure the bearing stress directly or modify the tests to induce bearing failures in order to make inferences about the relative performance of the various bar and hook geometries with respect to concrete bearing stresses.

On January 14th, I referenced Klein's curved bar node work. On September 13th, CEL referenced the same. I've included a snippet from that document below which, in my opinion, pretty much says it all with regard to how and "exiting tension" exacerbates the concrete bearing stresses inside the rebar corner.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Off topic post...

Discussions like these are why I cruise eng-tips. Seeing the heavyweights here discuss topics in more detail than a lot of us even knew existed.

I have now forwarded this thread to many people in our office when they ask about retaining wall stuff. I think it blows many of their minds.

Hey Kootk; My thanks. The compliment stands, just as a quick FYI...

As to the discussion at hand: I think you're conflating seismic and blast type work (ie: A proper frame, which can be closed by force that actually exists on both sides, and a simple anchoring situation, which is what I continue to believe exists here.

I may break out the hydrocode this weekend. You've gotten me very keen to know what an artificial physics environment might say about the stresses at the bar in a retaining wall. I suspect we may have just stumbled, or rather started to trample upon, something that our profession does not yet understand fully.

BUT: If you're right, why don't all those crappy, poorly detailed retaining walls fail? I think they don't fail because, while they are much poorer examples than would come from under your hand, they only really *require* the development of the bar to work in practice...

Quote (CELinO)

BUT: If you're right, why don't all those crappy, poorly detailed retaining walls fail?

1) In the wild, they probably don`t fail because of material and load factors and, in my estimation, the low probability of retaining walls actually seeing the design soil pressure. You know, the usual stuff that makes structural engineering an oddly consequence free space and makes it nearly impossible for clients to distinguish good design from bad. Additionally, retaining walls seem to usually experience overturning stability failures prior to material failures for whatever reason.

2) In the lab, the joints do fail prematurely. I posted a good deal of information on that above. Of course, if you don`t accept my arguments regarding the demands on the joints, you will also be unlikely to accept the relevance of the testing.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

Before proceeding further, I should mention that I only have access to a partial google doc on this: Link.

Yep, that's the one I was looking at.

Quote (KootK)

To me, it appears that the larger bars did not outperform the smaller bars.

Regarding the specimens without crossbars, NONE of the SMALL bars broke, MOST of the MEDIUM bars broke, and ALL of the LARGE bars broke.

Quote (KootK)

pretty much says it all with regard to how and "exiting tension" exacerbates the concrete bearing stresses inside the rebar corner.

Yeah but it doesn't work like that in reality. The bar isn't a rope draped over a pulley. It's a sticky bar bend around a sticky surface.

I see the research you've posted and it is interesting and compelling, but I do not believe it applies. Overall my position on this is becoming more entrenched and, I think, simpler...

Like a column that someone omits from a design calculation, but installs, which later results in a failure of a footing or beam below upon which it rests, you cannot decide you don't have a load path when you clearly do.

You want to treat the corner as a pulley, or at least as analogous to one... But like Tomfh has pointed out, it isn't a good comparison.

Once we have sufficient concrete around the full development length of the hook, the stresses have been transferred into the surrounding concrete, whether or not the bar then continues or does double duty in the footing. The load path is there, and the stress field will be created, unless you place bond breaking materials. That's why so many of those tests include bond breakers and jacketing tubes... Because they are only getting any result at all in their testing by forcing an artificial situation. The stresses actually transfer into the surrounding concrete, and then I believe they act as I outlined in my last sketch.

Quote (Tomfh)

Yeah but it doesn't work like that in reality. The bar isn't a rope draped over a pulley. It's a sticky bar bend around a sticky surface.

And that very "stickiness" is built into the STM model as circumferential bond force.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Fig 6 is all wrong.

If the applied load and the "exit" load are the same then why are all the internal forces totally assymetric?

Why is there a bond force at all if you're taking it all out via the exit bar?

Why does maximum normal stress occur at the start rather than at the apex?

And I still agree with you, in the case where this is a closing moment in a frame.... Just like the Fig 6. title "Unequal tie forces in a frame corner result in bond stress along the circumference of the bend."

So, I think you're still mixing up a force at each side, call it F, with a developed bar, AsFy. In the case of the closing frame moment, we have the two forces, equal and opposite, causing the closing moment. So all of our calcs refer to this as 2x(etc,etc) in terms of the resulting compressive stress. But where we are developing a bar into the footing, we have something more analogous to directly subjecting the inner bend to AsFy.

So: If we consider two forces, or one force closer to AsFy, I think the physics is the same, but the math looks different.

Quote (CELinO)

Once we have sufficient concrete around the full development length of the hook, the stresses have been transferred into the surrounding concrete, whether or not the bar then continues or does double duty in the footing. The load path is there, and the stress field will be created, unless you place bond breaking materials.

So... as long as we have developed bars surrounded by concrete, everything will just take care of itself? Seriously? Tell that to all the poor bastards who have been toiling away on the strut and tie method for the last fifty years for, what, their health?

Quote (CELinO)

you cannot decide you don't have a load path when you clearly do.

I have not ignored a load path here. Quite the opposite: I've gone to great pains to actually establish one that can be quantified. The last FBD diagram that you posted is one of the classic offenders with respect to retaining wall joint detailing. Designers think that they can provide a bottom mat independent of the stem bar horizontal projections and, somehow, that obviates the necessity for establishing a mechanism for transfer of moment from the footing to the wall.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (Tom)

If the applied load and the "exit" load are the same then why are all the internal forces totally assymetric?

Who says they're the same? Certainly not me.

Quote (KootK)

In the general case where entry and exit bar tension is unequal, some force transfer does occur via bond stress.

Quote (Tom)

Why is there a bond force at all if you're taking it all out via the exit bar?

Because, in the general case, you're not taking it all out at the exit bar.

Quote (Tom)

Why does maximum normal stress occur at the start rather than at the apex?

Could go either way depending on the relative proportions I would assume.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

So... as long as we have developed bars surrounded by concrete, everything will just take care of itself? Seriously? Tell that to all the poor bastards who have been toiling away on the strut and tie method for the last fifty years for, what, their health?

The strut and tie method pulls the same sleight of hand. It dissolves the big loads into little circumferential ring tension stresses and then forgets about them.

Sometimes it adds cross bars (e.g. bottle and fan reinforcement) where the trick fails and cracks appear, but it's still sneaking the tension into the concrete.

Quote (CELinO)

In the case of the closing frame moment, we have the two forces, equal and opposite, causing the closing moment.

In the case of a common retaining wall, moment must be transferred between the stem and the toe. That moment is equal and opposite and all that jazz... just like with frames.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

Who says they're the same? Certainly not me.

It assumes Asfy load at both bars.

Quote (KootK)

Because, in the general case, you're not taking it all out at the exit bar.

But you said the load doubles? I.e. the load, plus all of it again.

Quote (KootK)

Could go either way depending on the relative proportions I would assume.

You are missing the point. Look at the symmetry (or lack thereof!). If the applied load and exit load are both As.Fy then then the internal forces/vectors should be mirrored about the diagonal. But fig 6. shows one-way bond stress and a triangular normal force diagram!

Quote (Tomfh)

The strut and tie method pulls the same sleight of hand. It dissolves the big loads into little circumferential ring tension stresses and then forgets about them.

Not so with the cureved bar node method that has been presented above. The circumferential bond stresses make their way into the radial struts. That's why, at the end of the day, the radial struts usually converge somewhere other than that focal point of the circle.

In the Klein document, they cover a slab/wall closing joint as one of the examples. I find it to be quite analagous to the stem / footing joint under consideration here.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

But again we are conflating two situations... In the case of a seismic frame, for example a beam-column joint, all of those detailing issues are critical. But if we are looking at a uni-directional stress model, we are back to what worked for generations of engineers who did not design reinforced concrete for cyclical loading: Near face corner bars and other "sins", which have little effective harm for such simple load scenarios.

I get the feeling you're trying to fit the retaining wall reality to the STM theory. I think we need a theory which explains the reality, and for more than simply brushing aside the cases you don't like as only being okay because of safety factors and the cautious conservatism of our design codes.

Quote (KootK)

Not so with the cureved bar node method that has been presented above. The circumferential bond stresses make their way into the radial struts.

Bond stress = concrete in tension, which neither STM (or any other design method) properly resolves. We just palm it off onto the concrete knowing she's good for it.

Quote (Tomfh)

But you said the load doubles? I.e. the load, plus all of it again

Quote (Tomfh)

You are missing the point. Look at the symmetry (or lack thereof!). If the applied load and exit load are both As.Fy then then the internal forces/vectors should be mirrored about the diagonal. But fig 6. shows one-way bond stress and a triangular normal force diagram!

I don't know what to say Tom. I've already clarified, repeatedly and at length, my opinion that 2X is the upper limit and that there is no need for symmetry. Again, from many, many posts back:

Quote (KootK)

In the general case where entry and exit bar tension is unequal, some force transfer does occur via bond stress.

If you're bothered by the asymmetry, that's a Tom issue, not a KootK issue.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

"I find it to be quite analagous to the stem / footing joint under consideration here."

But it isn't. That's a corner, not a slab which is held down and in place away from the joint.

[sarcasm] Next you'll want to be developing bars down into a slab rather than epoxying post-fixed anchors for minor columns. [/sarcasm] There just isn't a need to project the bar further than a competent development where the concrete can take the stress of the loads and resist this as an anchoring beam. That's why I think your continual use of a frame corner is a bad comparison. The frame corner cannot do what the footing is doing...

Quote (KootK)

If you're bothered by the asymmetry, that's a Tom issue, not a KootK issue.

KootK, if the load in each bar is As.Fy (as the diagram assert for 45 degrees case) then the internal load vectors have to be symmetric. Think about it.

In particular it is a complete nonsense to say that normal force is maximum at the entry point.

On a lighter note - isn't it funny that we can design bridges and buildings and stuff but we can't even agree on what a stupid hook does!

Quote (Tomfh)

On a lighter note - isn't it funny that we can design bridges and buildings and stuff but we can't even agree on what a stupid hook does!

It is funny. And telling in my opinion. I think that it's partly a consequence of our over-reliance on computers. We can "design" a building in ETABS in a couple of hours but nobody can detail a joint to save their lives. Of course I do realized that, in your estimation, it's me who cannot design a joint. And that's fine.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Honestly, I think it just goes to show that the devil is in the details. How many time do we get a project to 90% complete but that last 10% seems to take 90% of the time.

Professional and Structural Engineer (ME, NH, MA)
American Concrete Industries
www.americanconcrete.com

Quote (TME)

Honestly, I think it just goes to show that the devil is in the details. How many time do we get a project to 90% complete but that last 10% seems to take 90% of the time.

100% of the time this is how it goes

Quote (KootK)

Of course I do realized that, in your estimation, it's me who cannot design a joint.

I think most of us here can design joints just fine. :)

It's how they really work that we disagree on. I remain fairly convinced a hook doesn't work like Fig.6, despite the fancy vectors and equations.

"Of course I do realized that, in your estimation, it's me who cannot design a joint. And that's fine."

Actually I think you can likely design just about any joint you care to, and would to a wonderful job... In my case I just think you're over-thinking this type of joint.

I would love to see some instrumented full scale testing results on retaining walls. I know RMC was doing a bunch of full scale testing on retaining walls some ten years ago, but I think the focus of their work was the pressures in the soil behind the wall. I'll have to do some digging.

But I still will detail the joints as per our discussion here, and don't like the idea that the hook is simply terminated. All types of anchors, as well as development of regular bars, must occur in zones which are not under tension, plastic hinging or similar.

It's just the physics I think we need to understand better...

A few things that I'd like to clarify here:

1) The testing that I've been reviewing has been overwhelmingly monotonic. The reported deficiencies, as far as I know, have nothing to do with blast, fatigue, seismic, or any other manner of cyclic loading.

2) With regard to what has worked for generations, I take that to be the CRSI diagonal bar detail. And, thankfully, that detail tests well. Unfortunately, it has become clear from this thread and my personal practice that many engineers are now omitting the diagonal bars. Based on the testing, that would seem to be a matter of some concern as what we're currently doing tests poorly and is, in fact, not what has worked well for us for generations.

3) Whenever folks don't like the results of my STM arguments, they're often quick to toss me into the "Basket of Deplorable STM Lovers". "Overthinking" as it were.

Firstly, I have not suggested that everyone run off and start designing these joints via STM. Above, I've explicitly indicated that even I don't do that. I've simply been doing what I usually do: using STM as a rational basis for discussing the mechanics of the situation.

Secondly, to my knowledge, STM is pretty much the only established tool that we have available to us for analyzing joints (disturbed regions) that have not been extensively tested. When I make use of STM to examing problems and get accused of "over thinking", what I hear in my head is "I don't like the answer so I'm just going to refuse to acknowledge the problem". STM doesn't need to be the method of design but, if an engineer can't find some way to justify a connection schematically via STM, I will always question whether or not that engineer has really identified a valid load path.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

There are thousands of retaining walls built of CMU, not one of which has anything approaching the diagonal bad shown in thr CRSI detail.

They too have been successfully used for generations without an abnormal number of failures.

I am a big fan of STM, and have used it in professional work. I just think you are overlooking confinement forces that assist. And a fully developed bar extending beyond into the footing is not the sinful and crap detail you've derided above despite my FBD being "one of the classic offenders with respect to retaining wall joint detailing."

Did you misunderstand what I showed? It was not simply a 90 deg hook developed into the footing, but a FBD of the standard detail used all over the world, not the junk detail where the bar never leaves the area under the stem.

Quote (CELinO)

There are thousands of retaining walls built of CMU, not one of which has anything approaching the diagonal bad shown in thr CRSI detail.

True. But then I would attribute the lack of failures to the same two factors that I listed previously for concrete walls: load and material factors and the low probability of the estimated soil pressures coming to pass. I would add to that the fact that we design CMU to relatively low stresses and, as such, we do not work block walls nearly so hard as we do concrete walls.

Quote (CELinO)

I just think you are overlooking confinement forces that assist.

Okay, so how does one go about using those confinement forces to quantitatively demonstrate that a joint is adequate when the only criterion met is bar development? As structural engineers, we're usually in the business of what can be demonstrated, not what can be qualitatively imagined.

Quote (CELinO)

Did you misunderstand what I showed? It was not simply a 90 deg hook developed into the footing, but a FBD of the standard detail used all over the world, not the junk detail where the bar never leaves the area under the stem.

Not sure. I'm familiar with five versions of your detail, as shown and listed below.

#1. Standard hook. Abomination. I see this in about 45% of cases where a detail like yours is used.

#2. Ld. Better but still a fail. The A & B bars need to be lapped, not just developed. I see this in about 10% of cases where a detail like your is used.

#3. Full tension lap. Close but not quite good enough. The lap needs to account for the lateral offset between the A & B bars (via non-contact lap or STM). I see this in about 5% of cases where a detail like yours is used.

#4. Maximum extension to end of toe. Adequate but probably by accident rather than understanding. At this point, there's really no need to have the separate bottom mat of reinforcement unless there is some oddball reason for it in the heel. We're just wasting material and telegraphing our ignorance. I see this in about 40% of cases where a detail like yours is used.

#5. All good. I've yet so see this once in practice.

Perhaps your are that one dude in all of the milky way who's doing a legitimate #5. You tell me.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Why is a hook an abomination if the bar embedment plus hook is sufficient to develop the bar?

Is there any evidence of standard hooks embedded 10 inches deep tearing out of retaining wall footings?

Quote (Tomfh)

Why is a hook an abomination if the bar embedment plus hook is sufficient to develop the bar?

At this point, I'll have to assume that you're either joking or trying to get me to embark upon a program of cutting / self-immolation.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Here Tomfh, read this excellent topic on eng-tips: http://www.eng-tips.com/viewthread.cfm?qid=401855

(Sorry, don't mean to tease but I couldn't resist)

Professional and Structural Engineer (ME, NH, MA)
American Concrete Industries
www.americanconcrete.com

Quote (KootK)

At this point, I'll have to assume that you're either joking

Im not joking. It's a genuine question.

If bars break before concrete breaks then what's the problem with that arrangement?

Quote (KootK)

The lap needs to account for the lateral offset between the A & B bars (via non-contact lap or STM)

Forgive me but shouldn't there be no effective difference in the lap-splice length for contact or non-contact lap splices? Why do we need to account for it here? Apparently my retaining wall joints fall in your 5%... So close.

Professional and Structural Engineer (ME, NH, MA)
American Concrete Industries
www.americanconcrete.com

Quote (TME)

Here Tomfh, read this excellent topic on eng-tips: http://www.eng-tips.com/viewthread.cfm?qid=401855

RDRR

I've looked over it. I still can't see what's so bad about a hook going into the bottom of the footing. KootKs fig 13-28 above calls it satisfactory.

Now, now... No pulling a Budhist Monk on us Kootk; You're well valued and liked around here... If people are asking questions it is to learn. I hope. And put away the razor blades; None of us are fourteen and in love for the first time anymore (lost a good friend to three years of psychiatric treatment to that one, so it takes a fair amount of strength to try to make a joke).

"Forgive me but shouldn't there be no effective difference in the lap-splice length for contact or non-contact lap splices?"

There is a major difference between non-contact and contact lap splices. The distance of separation must be added to the length of the lap splice. This is well treated in many codes, though I do not know the ACI code well enough to swear to it being addressed, it *is* well addressed by the ACI committee on lap splicing. See http://civilwares.free.fr/ACI/MCP04/408r_03.PDF

Back to the point at hand: The "all the way" detail is standard practice in New Zealand, and what I typically detail unless it is a particularly long toe in the footing. It is is particularly long, then I do No. 5. I'm not kidding, and can probably dig up three or four examples from my records if you want to see it done in practice.

Quick question: Do you really not like the steel extended to the heel? I actually like to include this as temperature and shrinkage, and just to keep the thing easy to construct. I would worry that if you curtail these bars, and the Contractor mixed things up, they might curtail the TOE bars! Take that plus an over-zealous pour, or a pre-pour rebar check by a green intern who misses the issue and blamo you've got a real problem...

WAIT... Just checked my records and I don't do five. I would have to call mine an option "6", where the full development length, including compensation for offset lap splice, starts from the toe-side of the stem to footing interface. So my lap is longer than you think it needs to be; You may continue to believe that no one does it your <<All good/Never seen in practice>> option. *friendly rib jab*

And while I have no reason for doing that beyond my hatred of development within a beam-column joint (which is likely a habit from all my seismic and blast work), I will keep doing it and much prefer to start my development lengths outside of anything that looks like a beam-column joint.

I am genuinely enjoying this discussion; I really hope it isn't coming at the expense of anyone's sanity or zen. *looks at KootneyKid semi-concerned*

And I still think our theories must match reality, not the other way around. I just don't buy that all these retaining walls are only surviving because of load factors and other conservatism. Even if that is true, it only means that we should be reducing our load factors in order to produce economical designs. We are in the business of efficient use of client and overall economic resources, not Pyramid building....

On a personal note, I think we may have found something I am sufficiently interested in to go back to Uni and research...

"I've looked over it. I still can't see what's so bad about a hook going into the bottom of the footing. KootKs fig 13-28 above calls it satisfactory."

13-28 shows the bar leaving the beam-column (like) joint and being developed into the toe, not just a plain 90 deg hook development. You can't reliably develop a 90 deg hook in a tension or high (unconfined) shear zone. That's why it needs to go through to the toe, where it will be in a mat of concrete under compression/subjected to the compression strut coming from the stem.

Quote (CELinO)

13-28 shows the bar leaving the beam-column (like) joint and being developed into the toe, not just a plain 90 deg hook development.

Yes, that's what I was referring to. Sorry for the confusion. The cog leaving into the toe so that the hook grabs onto nicely confined concrete. The other way isn't as efficient...

Personally I do #4 except with the steel going under the bottom mat.

Quote (CELinO)

And I still think our theories must match reality, not the other way around. I just don't buy that all these retaining walls are only surviving because of load factors and other conservatism.

I agree here. We engineers love to second guess reality. Either reality being wrong when something that the calcs say works doesn't end up working, or vice versa and saying something "should have failed".

Quote (CELinOttawa)

"Forgive me but shouldn't there be no effective difference in the lap-splice length for contact or non-contact lap splices?"

There is a major difference between non-contact and contact lap splices. The distance of separation must be added to the length of the lap splice. This is well treated in many codes, though I do not know the ACI code well enough to swear to it being addressed, it *is* well addressed by the ACI committee on lap splicing. See http://civilwares.free.fr/ACI/MCP04/408r_03.PDF

Hmmmm, I don't know much of the international codes so maybe this is a non-US thing? I'm almost 100% sure that ACI 318 does not require any special treatment of non-contact lap splices short of verifying that they're not spaced too far. Further, I didn't fully re-read ACI 408 but the the non-contact lap splice section in there doesn't appear to mention that lapped bar spacing needs to be considered and showed that the bond strength improved for non-contact lap splices.

Am I missing something?

Professional and Structural Engineer (ME, NH, MA)
American Concrete Industries
www.americanconcrete.com

Quote (TME)

Am I missing something?

Nope. I was a bit imprecise with my terminology. I consider there to be two types of non-contact lap splices per North American codes:

1) Non-contact laps between bars less than 6" apart in which case no addition need be made to the lap length.

2) Non-contact laps between bars greater than 6" apart in which lap lengths need to be extended via STM theory, methods found in foreign codes, or judgment.

So yeah, depending on your bar spacings, there are cases where regular tension laps would get the job done.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (CELinO)

Quick question: Do you really not like the steel extended to the heel?

Obviously, it's not a problem from a performance perspective. It makes me queasy on two fronts:

1) It seems to befuddle designers. Most of the time when I see the bottom mat, I also see the hook bar details that I dislike.

2) It strikes me as wasteful in many applications.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (CELinO)

And I still think our theories must match reality, not the other way around.

Quote (Tomfh)

I agree here. We engineers love to second guess reality. Either reality being wrong when something that the calcs say works doesn't end up working, or vice versa and saying something "should have failed".

I disagree with these sentiments strongly on a philosophical basis. We are applied scientists and the scientific method works like this:

1) Observe something.
2) Construct a theoretical model to explain what you observe.
3) Test your model in a controlled environment.
4) Confirm or refute your theoretical model.

To devalue the importance of the theoretical model is to throw true understanding out the window altogether.

Past history of performance without a testing validated model has value but is, without question, a weak form of understanding.

I do not accept that a retaining wall designed with safety factors and field tested against wildly unpredictable soil pressures constitutes "a controlled environment".

The testing on opening and closing joints has already been performed and some of it has been posted above. That is step three and it has definitively identified a problem with some of our details. The missing piece of the puzzle here seems to be that you guys refuse to accept the testing as relevant to retaining wall joints.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (CELinO)

You can't reliably develop a 90 deg hook in a tension or high (unconfined) shear zone. That's why it needs to go through to the toe, where it will be in a mat of concrete under compression/subjected to the compression strut coming from the stem.

I disagree with this. The whole point of a hooked anchorage is to quickly develop the tension potential of a reinforcing bar. It would be useless if it didn't apply in tension zones. Most exterior beam column joints have hooked development taking place withing a region of flexural tension and high shear. The reason that the bars need to go around the corner is because the tension force needs to go around the corner. And that's not the same thing as bar development.

Quote (Tomfh)

Why is a hook an abomination if the bar embedment plus hook is sufficient to develop the bar?

Quote (Tomfh)

Im not joking. It's a genuine question.

Quote (Tomfh)

Yes, that's what I was referring to. Sorry for the confusion. The cog leaving into the toe so that the hook grabs onto nicely confined concrete.

So we're in agreement then? #1 is a detailing abomination?

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

So we're in agreement then? #1 is a detailing abomination?

No, I don't agree with that. I agree with your link that it appears a satisfactory detail. Not necessarily the absolute best detail, but satisfactory. It's silly to call it "an abomination" when it can yield the bars and has proven itself the world over.

An "abomination" is detail posted above where the cog is barely embedded at all, and sitting in the very top of the footing.

Personally I do #4, except the cog goes under the bottom cross bars.

Quote (Tomf)

I agree with your link that it appears a satisfactory detail.

I believe that you are mistaken. There is no link of mine indicating that a standard hook on the stem verts is acceptable. Certainly, that is not what figure 13-28 indicates. Figure 13-28 shows the hooks being extended out to the toe.

Quote (Tomfh)

It's silly to call it "an abomination" when it can yield the bars and has proven itself the world over.

The details that have been proven the world over are the CRSI detail with the diagonal bars and, to a lesser extent, the #4 detail with the bars extended to the toe. What is silly, in my opinion, is confusing simple bar development with proper joint design. If all it took to design a joint were bar development, why has the world bothered with:

1) Developing the strut and tie method? Lot's of effort there.

2) Developing joint design guides like this and this.

Why publish hundreds of pages on joint design when it could just all be summed up with "develop the bars"? That would take what, a paragraph? Silly researchers and code committees...

Detail #1 has a version of the problem show below from this document by one of the gods of concrete design. And whether or not the hook sits above or below the bars make no appreciable difference. It's not as though reinforced concrete design is about hanging reinforcing bars from other reinforcing bars. Of course, it's the very same anchorage issue that I highlighted in the STM model that I posted back in March. It would be the concrete breakout potentially represented by those two struts coming up from the hook location.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

There is no link of mine indicating that a standard hook on the stem verts is acceptable. Certainly, that is not what figure 13-28 indicates.

It doesn't dimension it but it's pretty close to a standard hook. It's well short of your requirement "Ld + ???"

Quote (KootK)

Detail #1 has a version of the problem

A slab hanging from a beam is different problem. That being said, only Gilbert would think that slab is going to fall off that beam due to the slab diagonal struts not being anchored up to the top of the beam. I've always wondered that - why are STM proponents so happy to rely on unreinforced concrete to resolve the diagonal tension struts in regular slabs (and beams for that matter). If STM is the new state of the art, and we aren't to rely on unreinforced concrete surfaces, shouldn't we be scrapping all notions of inherent shear capacity? Ties throughout in all slabs from now on?

Quote (Tomfh)

It doesn't dimension it but it's pretty close to a standard hook. It's well short of your requirement "Ld + ???"

The intent seems to pretty clearly be extending the corner bars to the end of the toe. With that being the case, the additional bottom bars in the toe become extraneous and lapping requirements would be moot. The prime requirement would be successful anchorage beyond the face of the stem.

Quote (Tomfy)

A slab hanging from a beam is different problem.

It is a different problem but the detailing issue at play is the same: planes of weakness being created where concrete must resist tensile stress for a complete load path.

Quote (Tomfh)

That being said, only Gilbert would think that slab is going to fall off that beam due to the slab diagonal struts not being anchored up to the top of the beam.

While Gilbert's suggestion will lead to better detailing in my opinion, I agree that it shouldn't be strictly necessary. It's probably more important to address hanger bar issues with the stirrups.

Quote (Tomfh)

I've always wondered that - why are STM proponents so happy to rely on unreinforced concrete to resolve the diagonal tension struts in regular slabs (and beams for that matter). If STM is the new state of the art, and we aren't to rely on unreinforced concrete surfaces, shouldn't we be scrapping all notions of inherent shear capacity? Ties throughout in all slabs from now on?

Well, this STM proponent has no intention of abandoning the use of diagonal concrete tension to resist shear. It works and is the most economical solution in many instances. When you're in the Bernoulli region of a member, STM is quite unsuitable as a design tool and really morphs into just a framework for the discussion of how load moves around. To that end, a lot of folks will draw STM models using diagonal concrete tension shear resistance as "phantom" ties. That's what Teguci's done in his model below. Everywhere that you see the SQRT symbol over a tie, it's diagonal tension rather than reinforcing. Sadly, it took me a few months to realize that the SQRT symbol was chosen in homage to SQRT(f'c), the driving parameter behind diagonal tension resistance. Sometimes I'm a little slow on the uptake.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

The intent seems to pretty clearly be extending the corner bars to the end of the toe

It looks fairly close to a normal hook to me, i.e. much the same as your "abominable" detail. The intent behind it being that way makes no difference to how it is.

Quote (KootK)

planes of weakness being created where concrete must resist tensile stress for a complete load path.

Quote (KootK)

this STM proponent has no intention of abandoning the use of diagonal concrete tension to resist shear. It works and is the most economical solution

Why this double standard? Why is the former abominable and the latter perfectly reasonable?

"Am I missing something?" Yes; As the distance of the gap between the bars goes above between 3 to 6 bar diameters, you are not getting the assured stress field transfer you need. The ACI Code (which I openly stated I don't know) apparently treats this by eliminating what I call non-contact lap splices. A lap splice within 3 bar diameters *is* a contact lap splice according to the codes I am familiar with, and only those which need to be extended because of the gap are called "non-contact".

Like so:



Leading to code clauses like:



This is the effect that Kootk was talking about, so I presume he said that you hadn't missed something because he is actually familiar with US practice where I am not...

But my fault entirely; I linked the wrong ACI paper and now can't lay my hand on the correct one. *sigh*

"The whole point of a hooked anchorage is to quickly develop the tension potential of a reinforcing bar. It would be useless if it didn't apply in tension zones. Most exterior beam column joints have hooked development taking place withing a region of flexural tension and high shear. The reason that the bars need to go around the corner is because the tension force needs to go around the corner. And that's not the same thing as bar development."

No. The way a hook works requires either a compression field or a tension field provided by confinement. That's STM; you can't have it both ways.

Perhaps I could have been more clear, try this: "You can't reliably develop a 90 deg hook in an unreinforced tension or high (unconfined) shear zone." Both are for the same STM reasons; You have to have something to grab onto, and that actually needs to be a compression field, regardless of how it is provided.

I still strongly disagree with your idea that the same force goes around the corner through the bar in all situations. That's frame thinking and doesn't apply to mass concrete, deep beams, or retaining walls. If you have an alternate load path that will result in a lower total stress on the concrete, that is the equilibrium state the concrete and bars will seek.

Quote (CELinOttawa)

The ACI Code (which I openly stated I don't know) apparently treats this by eliminating what I call non-contact lap splices

Interesting, yes this nomenclature difference was something I wasn't aware of. I was of course aware that if you exceed the spacing set in codes then you had to consider the compression struts and the resulting increase in development length. In the ACI code this is only allowed if you go to a strut and tie model but there are no "stock" provisions which cover it.

Professional and Structural Engineer (ME, NH, MA)
American Concrete Industries
www.americanconcrete.com

Quote (Tomfh)

The intent behind it being that way makes no difference to how it is.

Huh? The intent behind the detail and the research doesn't affect the conclusions? I'm afraid that I don't follow.

Quote (Tomfh)

Why this double standard? Why is the former abominable and the latter perfectly reasonable?

Firstly, one way shear has been shown to work reliably whereas breakout of concrete anchors (rebar in this case) has not. Are you familiar with the much maligned ACI Appendix D used in the US? It exists because of this problem. The Wheeler research that you keep coming back to intentionally created an environment where concrete breakout would be precluded: mass concrete and stirrups at 125 o/c. As such, that research is not relevant to the concrete in tension issues that are at play here.

Secondly, STM is just one possible method of analysis. Nothing about STM precludes the use of alternate methods that may utilize concrete in tension, particularly in Bernoulli regions. Same goes for folks that are proponents of STM. Being an STM fan doesn't mean that I deny diagonal tension shear resistance.

Quote (Tomfh)

It looks fairly close to a normal hook to me, i.e. much the same as your "abominable" detail.

Really? Perhaps look a little closer then. Like most laboratory work, the testing was done on tiny little bars for which a standard hook would be on the order of 120 mm. The tested bars extended out to the end of the toe creating hook extensions ranging from 400 mm to 600 mm.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (CELinO)

This is the effect that Kootk was talking about, so I presume he said that you hadn't missed something because he is actually familiar with US practice where I am not.

Exactly. Our Canadian concrete code is identical of course (below). The only thing that TME may have been missing is that, where the space between lapped bars would exceed 6" or 1/5 lap, a lap extension would be the order of the day. For the bars to be offset that far, spacing would need to exceed 12" which would be rare in my experience.

Quote (CELinO)

I still strongly disagree with your idea that the same force goes around the corner through the bar in all situations. That's frame thinking and doesn't apply to mass concrete, deep beams, or retaining walls.

Oh my. I can see now why we're having trouble finding common ground. The retaining wall joints is absolutely a frame joint and I encourage you to revisit your stance on that. Do you have a access to any of the CRSI handbooks? The shabby sketch that I posted on Jan 14, and have repeated below, illustrates the idea.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (CEL)

No. The way a hook works requires either a compression field or a tension field provided by confinement. That's STM; you can't have it both ways.

Perhaps I could have been more clear, try this: "You can't reliably develop a 90 deg hook in an unreinforced tension or high (unconfined) shear zone." Both are for the same STM reasons; You have to have something to grab onto, and that actually needs to be a compression field, regardless of how it is provided.

I was confused about your intent. I thought that you meant that you couldn't have a hook in a flexural tension zone. I agree, a hook ought to grab on to a diagonal compression strut. That's a big part of what lead to my confusion. You indicated that you didn't think that the concrete directly below the stem was a good place for a hook. In fact, that is the location where you would have the largest possible compression strut/field available to restrain such a hook. Using corner bars to restrain that monster strut, with properly anchored vertical and horizontal legs, is really the crux of this thread.





I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

I agree with KootK, the joint from wall base to stem is 100% moment connection joint. Here's a nice little figure from the CRSI retaining wall manual KootK referred to that shows the flexural forces going into this joint:



This is what KootK is showing with his curved bar node sketch. If possible I highly recommend picking up a copy of the CRSI retaining wall manual (ugh, all broken up into individual manuals now), lots of good info in there.

Professional and Structural Engineer (ME, NH, MA)
American Concrete Industries
www.americanconcrete.com

• http://files.engineering.com/getfile.aspx?folder=89265b56-da66-4ccb-85b4-dd

Quote (KootK)

Huh? The intent behind the detail and the research doesn't affect the conclusions?

What I meant was the final geometry and final performance is what it is, and isn't affected by one's intent. What matters is the actual arrangement, and that arrangement with a relatively short extension does not appear to satisfy your critical requirement of Ld + ???. Or does it? If it does satisfy your rules I am happy to stand corrected.

Quote (KootK)

Firstly, one way shear has been shown to work reliably whereas breakout of concrete anchors (rebar in this case) has not.

But where are all these cases of it not working the way we see epoxy bars not work? Why your need to explain away the lack of failures?

If it doesn't work in reality I want to know.

Quote (KootK)

Like most laboratory work, the testing was done on tiny little bars for which a standard hook would be on the order of 120 mm.

Can you provide this research. I am curious to see what they actually did.

Quote (TME)

Here's a nice little figure from the CRSI retaining wall manual KootK referred to that shows the flexural forces going into this joint:

Gee that bottom bar's well-anchored isn't it...

Quote (Tomfh)

What matters is the actual arrangement, and that arrangement with a relatively short extension does not appear to satisfy your critical requirement of Ld + ???. Or does it? If it does satisfy your rules I am happy to stand corrected.

It does satisfy my "rules". As I mentioned previously,

Quote (KootK)

With that being the case, the additional bottom bars in the toe become extraneous and lapping requirements would be moot. The prime requirement would be successful anchorage beyond the face of the stem.

Quote (Tomfh)

But where are all these cases of it not working the way we see epoxy bars not work?

In the lab, underpinning all the research on joints and anchorage breakout? I don't know man. I'm a practicing engineer, not a curator of the condition of the world's retaining walls.

Quote (Tomfh)

Why your need to explain away the lack of failures?

Uh... because you guys keep asking me to explain away the lack of failures.

Quote (Tomfh)

Can you provide this research. I am curious to see what they actually did.

I cannot. A previous employer had a hard copy; I've never been able to procure a PDF. If you're sufficiently motivated to track down a copy, the source document seems to be this:

Nilsson, I. H. E., and Losberg, A.
“Reinforced Concrete Corners and Joints Subjected to Bending Moment,”
Journal of the Structural Division, ASCE, V. 102, No.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

With that being the case, the additional bottom bars in the toe become extraneous and lapping requirements would be moot

So because it's too short to satisfy your requirements for (Ld + ???) therefore it does actually satisfy your requirement?!?

Quote (KootLK)

Uh... because you guys keep asking me to explain away the lack of failures.

The intention behind pointing out the lack of failures was to illustrate real world performance (or at the very least our impression of it), not to give you something to explain away.

Quote (KootK)

In the lab, underpinning all the research on joints and anchorage breakout?

If you can supply "all the research" showing that regular hooks in the toes have a bad habit of tearing out, I am more than happy to stand corrected and accept it as a detailing abomination.

KootK...If you're interested you can obtain a scanned copy of the Nilsson article through the Linda Hall Library (Link).

You know those bars that I labelled A & B in my panoply of CEL details above? It works like this:

1) Ld + ??? is about lap splicing the A & B bars.

2) When the B bars extend all the way, the A bars become unnecessary as the B bars cover the flexure every damn place.

3) With the A bars being unnecessary, the lap requirement becomes moot.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

3) With the A bars being unnecessary, the lap requirement becomes moot.

So the anchorage of B stops working once you add the A bars?

Quote (Deker)

KootK...If you're interested you can obtain a scanned copy of the Nilsson article through the Linda Hall Library (Link).

Thanks so much Deker. I just might given that:

1) I appear to be doomed, like Sisyphus, to have to defend my position against all comers in perpetuity and;

2) This thread is shaping up to be my personal magnum opus to be left behind to humanity when I pass.

I'm hoping to eventually acquire verb status. Something like: Dude! You really KootK'd the #@%$ outta that joint! I don't even care if it's meant as an aspersion, I just want the notoriety.



I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

I appear to be doomed, like Sisyphus, to have to defend my position against all comers in perpetuity

Sisyphus got up the hill.

Show me test results that show a standard hook going into the toe tears up the concrete and I'll happily concede it's an abominable detail.

Quote (Tomfh)

So the anchorage of B stops working once you add the A bars?

Nope. Rather, once the B bars are extended all the way, the A bars are no longer required. At that point, all that matters is the proper anchorage -- not lapping -- of the B bars.

Very much like this:

Quote (KootK - Earlier Today)

The intent seems to pretty clearly be extending the corner bars to the end of the toe. With that being the case, the additional bottom bars in the toe become extraneous and lapping requirements would be moot. The prime requirement would be successful anchorage beyond the face of the stem.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (Tomfh)

Show me test results that show a standard hook going into the toe tears up the concrete and I'll happily concede it's an abominable detail.

Well golly then, let me just go fire up the retaining wall yielder that I keep in the back yard for just such happiness emergencies.

Some details don`t get much research attention because the engineering community deems them so unworthy as to not justify the effort. Some other conditions that you won`t find much testing data for:

1) Effectiveness of rebar installed in the project down the street.
2) Effectiveness of rebar installed via teleportation fifty years in the future.
3) Effectiveness of rebar that`s been replaced with black licorice.





I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

That last batch was for opening joints. As in the connection between the heel and the wall. It would be helped some by the presence of the toe of course.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

Well golly then, let me just go fire up the retaining wall yielder that I keep in the back yard for just such happiness emergencies.

I was not asking you to go test retaining walls. You said there is "all the research". I was simply asking for you to show the research that shows a hook turning into the toe is no damned good.

Quote (KootK)

That last batch was for opening joints.

We are talking about cog going into the toe of a retaining wall, not opening joints with cogs turning outwards.

Quote (Tomfh)

We are talking about cog going into the toe of a retaining wall, not opening joints with cogs turning outwards.

I would argue that detail A in my last batch comes pretty close to proving my point about the abominable detail. Based on what you`ve told me of your understanding, it goes like this:

If the bars coming into a moment joint are developed with standard hooks, then the joint should be adequate.

By that logic, detail A should work. After all, the only reason to care about the direction of the hook is if you`re like me and you mistakenly think that you need to transfer rebar tension around the corner and into the to bottom of the toe.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Seriously Kootk? People can disagree with you without it being an all or nothing...

1. I think the orientation of the hook matters.
2. I believe it should be developed outside the stem.
3. I do not believe the joint behaviour is analogous to a beam column joint as the loading conditions are significantly different and the state of stress within the joint differ significantly.
4. I think the number of crappily detailed walls which never fail, nor even exceed SLS for thst matter, not to mention the CMU walls for which you disaprove of the detailling prove that the joint in question cannot be as easy to fail as you think. I believe it prooves that we do not need to get the load around the corner, but instead just need to ensure thst the area stressed by the bar development can handle the stress required.

Overall I suspect we're just going to have to agree to disagree, but I am hoping we can all still do that respectfully.

Quote (KootK)

I would argue that detail A in my last batch comes pretty close to proving my point about the abominable detail.

It's a completely different situation. It's an opening joint with the cog going outwards. We are talking about retaining wall with the cog going into the toe.

Attacking the easy target that is detail (a) and declaring yourself victorious is pure strawman arguing.


You correct though that it shows we shouldnt automatically assume all embedded hooks can develop the bar. Some clearly cannot.

Quote (CELinO)

Seriously Kootk? People can disagree with you without it being an all or nothing...

What exactly are you objecting to here CEL? I wouldn't continue to debate with you if you weren't continuing to debate with me. It's kind of of a two way thing. And you'd led me to believe that you were enjoying the debate. That said, if you can't take the heat...

I believe that you are harboring an egregiously erroneous misunderstanding about the nature of these joints and I've very respectfully suggested that you reconsider you position. That's all. It's hypocrisy to jump into a nine month old thread, suggest that all my work is incorrect, and then complain when I defend my position.

Quote (Tomfh)

Attacking the easy target that is detail (a) and declaring yourself victorious is pure strawman arguing.

I didn't attack anything. I presented an argument that I believe in as an attempt to persuade. And no where -- not one damn place -- did I claim to be victorious.

Quote (Tomfh)

You correct though that it shows we shouldnt automatically assume all embedded hooks can develop the bar. Some clearly cannot.

All of the details can develop the bar. They fail because modes other than development kick in. That's the whole point. There's more to a joint than just development.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Turning into a mass-debate without a happy ending...

...wow, I got a lot of reading to catch up on this thread.

I've ponied up for the Linda Hall Docserve request on the Nilsson stuff. I'm going to hang back on my maniacal quest for world domination until I can report back with those results. Probably late next week. Rarrr!!!

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

I didn't attack anything. I presented an argument that I believe in as an attempt to persuade. And no where -- not one damn place -- did I claim to be victorious.

You referred to those details as "crap" and you said it "proving your point"

Quote (KootK)

All of the details can develop the bar.

No, they can't. The concrete ruptures first.

Quote (KootK)

I've ponied up for the Linda Hall Docserve request on the Nilsson stuff.

Good job. Please post the article when you get it.

What do I object to? Posts like 30 Sep 16 00:18. It is exactly that type of non-collegial and argumentative post that had me so sick of Eng-Tips I didn’t log in for the better part of a year.

Quote (Tomfh)

No, they can't. The concrete ruptures first.

Just because the concrete failed doesn't mean it wasn't developed. By my understanding you can "develop" a bar in a concrete cylinder barely large than the bar. Assuming tension only; at the location this bar terminates the concrete will fail well below the strength of the bar. Developed bar does not equal concrete sufficient to avoid failure; it's simply a bond strength check. Otherwise why would appendix D even exist?

Professional and Structural Engineer (ME, NH, MA)
American Concrete Industries
www.americanconcrete.com

Quote (TME)

Just because the concrete failed doesn't mean it wasn't developed.

We're clearly all at cross purposes now as it makes no sense to me to say a bar is being developed despite the concrete failing well before the bar breaks. Concrete breaking at 50% of bar break load = bar not developed.

Concrete failing and the bar pulling out at 50% of the load at which the bar breaks is about as clear a case of a bar not being developed as I can imagine.

Quote (TME)

Developed bar does not equal concrete sufficient to avoid failure

that precisely what I understand bar development to be - sufficient enough concrete that the bar breaks first. Sufficient enough concrete that the ring tension stresses around the bar are sufficiently diluted to avoid concrete rupture. It's better in the toe as the clamping force fights against this. When it goes the other way, into the heel, there's already tension and cracking to start with and so the hooked bar pullout stresses overwhelm the concrete a lot easier. Hence the opening joint with outward hooks being so "crap".

Quote (TME)

By my understanding you can "develop" a bar in a concrete cylinder barely large than the bar.

you need sufficient hoop-stress/ring-tensile capacity in the concrete around the bar to resolve the outwards strutting-forces/shear from the bar. A very thin concrete sleeve around the bar has very limited ring tensile stress. This is why concrete cover is so critical to bar development. It governs the thickness of concrete in ring tension around the bar.

Quote (Tomfh)

Concrete failing and the bar pulling out at 50% of the load at which the bar breaks is about as clear a case of a bar not being developed as I can imagine.

Yes, this is a lack of development if the bar is "pulling out". I'm talking about a failure of the concrete section:



Here I've provided the full development length of the bar and thus "developed" the strength of the bar; yet the concrete still failed. Perhaps I'm just using a different definition of "development length" as I think I get what you're saying? I just think "development length" is the wrong term for it.

Per ACI 318: "Development Length - length of embedded reinforcement...required to develop the design strength of reinforcement at a critical section."

Professional and Structural Engineer (ME, NH, MA)
American Concrete Industries
www.americanconcrete.com

Wow - What exactly constituted a "Development Length" seems to vary by more than what non-contact means in reinforced concrete.

So fsr I have found (paraphrasing):

1. Sufficient length of embedded bar to transfer all stress to the surrounding concrete without failure.
2. Sufficient length of embedment to prevent pull out.
3. Sufficient length to ensure the bar yields rather than causing a concrete breakout cone.
4. The length of the bar required after a hook. (NOT kidding - uck!)
5. A length dependant on bar size and concrete strength required for continuity of concrete construction.
6. Sufficient lenght to ensure the bar does not pull out more than X under Y percent of ultimate strength loading.

That was a couple of minutes with google. I think the issue is between two fundamental schools of thought:

A. Ld means the bar transfers all stress to the surrounding concrete without a failure.
B. Ld means the bar performs to some lesser standard where concrete is allowed to fail, something more akin to a Capacity Design approach.

Could it be that most codes mean 'B', while most engineers think 'A'? If so, then I am about to learn something very fundamental about concrete detailling that I have misunderstood as not being the case for non-seismic regions...

Quote (CEL)

So fsr I have found (paraphrasing):

1. Sufficient length of embedded bar to transfer all stress to the surrounding concrete without failure.
2. Sufficient length....

Yeah, this is what I think I going on here. To be clear I'm not disagreeing or agreeing with Tomfh regarding the wall details; I just wanted to make sure either I or he wasn't misunderstanding the term development length. From your search it does appear that some variations in what people mean with "development length" do indeed exist.

Quote (CEL)

Could it be that most codes mean 'B', while most engineers think 'A'?

I believe so; as I said above, why else would ACI 318 appendix D even exist?

Professional and Structural Engineer (ME, NH, MA)
American Concrete Industries
www.americanconcrete.com

Quote (Tomfh)

Ok, 1+1, can we agree it equals 2?

Well, I think 1+1 = 3 for sufficiently large enough values of 1.

Professional and Structural Engineer (ME, NH, MA)
American Concrete Industries
www.americanconcrete.com

Quote (CEL)

Wow - What exactly constituted a "Development Length" seems to vary by more than what non-contact means in reinforced concrete.

I understand it as the length of bar embedment beyond a cross-section required to yield the bar. If instead of yielding the bar is instead tearing out of the concrete (and resulting in joint efficiencies of 30-50%) then by definition it's not developed.

Aren't Appendix D and Anned D specific to post-installation?

I don't think they really justify anything about cast in place bar development. They are an entirely different animal and more about how drilling into set and cured concrete behaves...

Appendix D covers both cast-in dowels and post-installed. You're correct (at least in my opinion) that it's a little apples and oranges, but my point was that if concrete failure wasn't an issue then why wouldn't we just be calculating the "development length" for a dowel and calling it a day? Perhaps a little too apples and oranges but hopefully you get what I was alluding to.

Professional and Structural Engineer (ME, NH, MA)
American Concrete Industries
www.americanconcrete.com

Quote (CELinO)

What do I object to? Posts like 30 Sep 16 00:18. It is exactly that type of non-collegial and argumentative post that had me so sick of Eng-Tips I didn’t log in for the better part of a year.

Bah! That was just a little snarky humor. If Tom had to take the day off to cry in his bear and re-watch Bridgette Jones, he should let me now. I'll send a basket.

I well remember the thread the precipitated your exit: Link. And I felt badly about it once I realized that you were gone.

Despite your claiming the contrary, I believe that the truth is that you do not enjoy vigorous debates when they are with me. And I don't think that "why" really matters. Our styles are just incongruous I guess. Different strokes for different folks. What I don't understand is why you keep engaging me in these vigorous debates when they clearly make you miserable.

I was very happy to see you back here again after your long absence. And, in threads where my level of investment is much less, I've made a concerted effort to give you some space so that we can avoid the kind of arguments that led to your prolonged absence. To some extent, it will take two however. You'll need to stop "poking the bear" as it were.

So here's what I'm asking... begging of you. When you start getting frustrated, disengage. I tend not to get frustrated with these kinds of discourse so it's very difficult for me to tell when enough is enough. Perhaps I'm a high functioning autistic.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (TME)

Developed bar does not equal concrete sufficient to avoid failure; it's simply a bond strength check. Otherwise why would appendix D even exist?

TME is spot on with this. It's also an oddly contentious issue and has been debated at length here: Link. In fact, I do believe that TME may have stolen my favorite, snarky illustration of the concept from there.

Whenever I pitch the concept, folks jump all over me saying that it's too ridiculous to be relevant. I believe that the juxtaposition of its ridiculousness with its technical correctness is what makes the sketch so salient. It is a properly developed bar by all measures. And it's tension capacity is zilch due to concrete breakout issues.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

cry in his bear

Quote (KootK)

I well remember the thread the precipitated your exit: Link.

Hmmmm, I seem to be unintentionally dragging this thread towards that thread by discussing the definition of development length. I'll leave that debate for another thread lest we make this thread any more contentious.

Quote (KootK)

I do believe that TME may have stolen my favorite, snarky illustration of the concept from there.

Guilty as charged, it was a good example.

Professional and Structural Engineer (ME, NH, MA)
American Concrete Industries
www.americanconcrete.com

And I find it most ironic, and funny, that I see the two threads as having now become nearly interchangeable... Only this time you've gotten all the stars, and last time it was more even keel.

I have become quite annoyed at the way that Eng-Tips becomes a "I'll vote my view up" game instead of a "Hey, that was genuinely helpful; I'll give them a star to thank them for their effort", which is what I think was/is intended.

As to the issues with Development Length: I have been, and as best as I can tell always will be, a strong believer in the reliability of a developed bar to pass the subject stress into the surrounding concrete. After this has occurred, you must then resolve the stress elsewhere. That's good detailing practice, and effectively what people do (without the thinking that I believe should go into our work) when they insist on developing a bar which has already been developed (ie: Pass a "tension" around a corner and anchor elsewhere).

Look: Everything we do it based on a gross number of assumptions. Which way you design your reinforced concrete *should* vary based on how you choose to handle the design. I suppose I am a "Bernoulli Region" thinker, and you have clearly become an avowed Strut-and-Tie proponent. I prefer modified compression stress field in my work, so I don't really think about STM very often anymore, other than for pile caps and inverted corbels carrying an existing structure on a new exterior pile.

And you're right. That thread basically left me entirely sickened with this site, like this one has nearly now managed to do. I wish I had the spare time to run an Ansys Mechanical simulation on this question, but I'm just too busy. My only hope is that you realize that I believe you're right about the assumptions required for your work, that they don't always apply to how someone else is choosing to do a design, and further that NEITHER ONE is a true reflection of what is going on, but a practical model with necessary simplifications to enable us to get some stuff done in the real world.

Also: Canada, and even more so the USA, have become a "There is one right way to design X". That's cook-book thinking and leads to crappy, overly conservative designs and chokes out innovation, and particularly to a poor quality of University graduate from the Engineering schools. I have worked with new grad interns from universities in five countries so far, and we don't compare well against the skills coming out of the NZ universities.

I hope and pray for the day where Canada adopts a genuinely Performance-Based approach and the codes loosen up and allow for practical thinking again. That and I would welcome Authorities Having Jurisdiction requiring the submission of calculations with permits. The fees are so low sometimes that I find it hard to believe anyone is doing more than simply drawing lines, labeling best-bet beam sizes from a half-checked model and submitting the work.

If it isn't already obvious from the discussion (above), I do very little work with A23 and have not become reacquainted with the finer points of detailing of Reinforced Concrete according to the Canadian Code, despite being back designing in Canada. I simply don't do reinforced concrete outside of my very specialized position with the Federal Government, where we use a custom code for high strain rate and impulsive loads. I'll let you fill in the blanks on that one, but suffice it to say that the real world behaviour of concrete far outweighs the belts-and-suspenders while reading a cookbook approach that our code demands.

I hope that my next job for design in reinforced concrete in Canada is a tilt-up job. I would love a chance to bring some of the practical detailing innovations from New Zealand (which would satisfy the A23 code) to local practice, and I really need such a job to refamiliarize myself with the Canadian code. Knowing NZ 3101 like the back of my hand and being functionally familiar with the codes for Hong Kong, Vietnam, and French Polynesia (specifically New Caledonia) isn't really what I need for local practice. I just haven't been willing to get to studying the new code until I have a job that needs it; Only time will tell if I get the chance.

In the meantime, I think I'm out.

While I'm anxious to get to "done" as well, I feel that I've come up with one more salient argument that hasn't yet seen the light of day.

Below, I've included a blurb from MacGregor's text which references a related ACI clause. Essentially, it's a fundamental statement of what reinforced concrete is. Not just concrete mind you. REINFORCED concrete. The fundamental principle being that, in reinforced concrete, flexural tension is passed from reinforcing bar to reinforcing bar without utilizing concrete in tension other than, perhaps very locally at lap splices. It's an important distinction because, as has been pointed out, concrete does have some tension capacity that can be mobilized in certain cases (app D etc).

Here's a summary of my thinking with this:

1) I went with an L-shaped retaining wall because getting rid of the heel simplifies things in a helpful way.

2) If all that matters is development, then I would argue that cases 2 & 3 are effectively identical.

3) Case 1 would pass the RC concrete litmus test indicated in the blurb at section A-A.

4) Neither case 2 nor case 3 would pass the RC concrete litmus test indicated in the blurb at Section A-A.

5) While cases 2&3 would have some capacity as a result of concrete tensile stress, that capacity would be greatly reduced.

6) I've never encountered a design method that relied on concrete in tension at the moment joint.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (CELinO)

The fees are so low sometimes that I find it hard to believe anyone is doing more than simply drawing lines, labeling best-bet beam sizes from a half-checked model and submitting the work.

Based on my time in engineering management, this is absolutely true. It was great fun to manage a stable of poor bastards who didn't have the budget available to do tight technical work but then had to deal with a KootK review at the end of the line.

And your statement below is a big part of why. Frankly, I'm ashamed of Canada on this front. What did we think would happen in a competitive environment sans sheep dog? First, folks get sloppy in order to make more money. Second, folks stop having the fee available to do anything but be sloppy. Not. Impressed. With selves.

Quote (CELinO)

That and I would welcome Authorities Having Jurisdiction requiring the submission of calculations with permits.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

I feel that I've come up with one more salient argument that hasn't yet seen the light of day.

I'm glad you did; that post summarizes in a very concise manner exactly why I think you're right. If people can't agree on what you've shown there, then it's probably best to just agree to disagree and move on.

Professional and Structural Engineer (ME, NH, MA)
American Concrete Industries
www.americanconcrete.com

So here's the Nilsson stuff which I'll let folks interpret for themselves. Really no new revelations beyond what has already been covered except, perhaps, the importance of the diagonal bars to crack control. There was no testing specifically for retaining walls having only standard hooks at the bottom of the stem bars. Reinforcing was 10M. Stems were 200 wide where as I estimated 250.

I won't post the entire article because I harbor some sporadic and selective respect for intellectual property rights. That said, if anyone has any specific questions, I'm happy to review the article and post whatever answers I can.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

For what its worth, U70 is what I typical see in CA and what we use almost all the time. Just wish they would only use real units, in, ft. etc.

Quote (TME)

Developed bar does not equal concrete sufficient to avoid failure; it's simply a bond strength check.

Quote (KootK)

TME is spot on with this.

Quote (KootK)

The fundamental principle being that, in reinforced concrete, flexural tension is passed from reinforcing bar to reinforcing bar without utilizing concrete in tension other than, perhaps very locally at lap splices.

Fundamentally incorrect. Stress development IS concrete tension. Specifically it is perpendicular concrete tension along the lengths of the bar. It's not simply some magical "bond" between the bar and concrete which is unrelated to the concrete tensile-rupture. Read the Gilbert link KootK posted about. As gilbert puts it - "bar anchorage is concrete tension". That's what it is.

A straight bar embedded in slab or tube of concrete (TME and KootK sketch) is irrelevant to our problem because the applied actions and failure plane are perpendicular the anchorage tensile forces meaning you can forget about them.

In our cases with hooks into a heel the anchorage tensile forces align with other tensile forces, and thus it's now critical (hence the poor joint efficiency). The concrete is preloaded with tension and hence the bars cannot develop. They can no longer induce the tensile forces needed to develop the bar because the concrete tensile capacity is almost used up already in those areas. As CELinO points out, developement is much worse in tensile zone. It is much harder to develop the bar in these tensile zones. In the case with the hook turning into the toe there is great confinement so anchorage is much better.

This idea that development is merely a bond force which is unrelated to rupturing of the surrounding concrete is ridiculous. It is simply ridiculous to call bars which are tearing up concrete as being "developed bars". A developed bar is one that DOES NOT rip up the concrete before it yields. It is a bar that does not overwhelm the concrete's tensile capacity prior to it yielding.

Quote (Tomfh)

Read the Gilbert link KootK posted about. As Gilbert puts it - "bar anchorage is concrete tension". That's what it is.

Gilbert's just talking about the perpendicular splitting forces that develop along bars as they anchor, slice, and develop. That's something that I've acknowledged previously and simply isn't part of the primary structural action.

Quote (KootK)

flexural tension is passed from reinforcing bar to reinforcing bar without utilizing concrete in tension other than, perhaps very locally at lap splices.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (Tomfh)

It is simply ridiculous to call bars which are tearing up concrete as being "developed bars". A developed bar is one that DOES NOT rip up the concrete before it yields. It is a bar that does not overwhelm the concrete's tensile capacity prior to it yielding.

Using ACI appendix D, I ran an analytical experiment. I put it together quickly so, no doubt, there's a mistake in there someplace. I'm sure that someone will point it out for me in short order. For now, here are the details:

1) Went with #8 bars at 12" o/c as a representative retaining wall case.

2) Calculated the tension demand along the bars in kip/ft.

3) Calculated the resisting concrete breakout capacity, also in kip/ft.

4) The capacity in this instance is less than half of the demand.

5) I used phi = 0.9 on the bars. To rupture a bar, phi = 1.25 is probably more apt.

6) It plays out better for smaller bars and worse for larger.

If you do this same exercise for a single bar, all by it's lonesome, you usually will have enough breakout capacity. That's not because development guarantees breakout capacity however. Rather, it's just how the numbers pan out. When edge or group conditions apply, development often will not guarantee adequate breakout capacity, as seems to be the case in this instance.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

Gilbert's just talking about the perpendicular splitting forces that develop along bars as they anchor, slice, and develop.

Quote (KootK)

That's something that I've acknowledged previously and simply isn't part of the primary structural action.

I don't agree you've acknowledged it because you are characterizing the tension as "just a perpendicular splitting force" and are saying it's has nothing to with overall failure.


The perpendicular tension is sometimes related to primary failure (i.e. global forces in the concrete), hence it being more difficult to develop a bar through a tension field. Hence the hook in detail (a) being far less developed than the hook in your "abominable" detail. In detail (a) the perpendicular splitting forces are pushing against concrete that's already wanting to pop out, and the joint fails early. In the abominable detail the hook wraps back into the confined core, where the concrete isn't already at breaking point.

Your rod-snapping and frustrum-pull-out examples do not have global tension forces aligned with the bar perpendicular splitting forces, plus the geometry ensures the concrete wont split along the bar, so it's cheating to cite these as proof that splitting along a bar is entirely unrelated to so-called "primary failure".

Quote (Tomfh)

The perpendicular tension is sometimes related to primary failure (i.e. global forces in the concrete), hence it being more difficult to develop a bar through a tension field. Hence the hook in detail (a) being far less developed than the hook in your "abominable" detail. In detail (a) the perpendicular splitting forces are pushing against concrete that's already wanting to pop out, and the joint fails early. In the abominable detail the hook wraps back into the confined core, where the concrete isn't already at breaking point.

I'm having a hard time parsing out what you're trying to get at here. How about a sketch or two?

Quote (Tomfh)

Your rod-snapping and frustrum-pull-out examples do not have global tension forces aligned with the bar perpendicular splitting forces, plus the geometry ensures the concrete wont split along the bar, so it's cheating to cite these as proof that splitting along a bar is entirely unrelated to so-called "primary failure".

Cheating huh? I thought that one of your arguments was that developed bars can't be pulled out of concrete by way of concrete breakout. I've show an analytical example of that assumption being inaccurate. Seems perfectly valid to me.

Quote (Tomfh)

This idea that development is merely a bond force which is unrelated to rupturing of the surrounding concrete is ridiculous. It is simply ridiculous to call bars which are tearing up concrete as being "developed bars". A developed bar is one that DOES NOT rip up the concrete before it yields. It is a bar that does not overwhelm the concrete's tensile capacity prior to it yielding.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

I'm having a hard time parsing out what you're trying to get at here. How about a sketch or two?

What I'm getting at is that anchorage tensile forces aren't just "perpendicular splitting forces" that are unrelated to overall failure. Yes in most occasions these perpendicular splitting forces do not interact critically with the concrete tension forces (e.g. your go to example above), but in many occasions they do, and failure occurs. As CELinO puts it, the anchorage needs to grab onto something. If that something (concrete tension capacity) is already used up then you have problems. here are a couple of examples. one a regular beam, and the second our joint cases.

This thread has been going on for 10 months and makes my head hurt.

SteelPE gets a star for the first non-partisan post in months... And because HYPOCRISY!!!! Hahahaha.... And manic tiredness of child-rearing. *sigh*

Thanks for the sketches Tom -- definitely food for thought.

Quote (Tomfh)

yes in most occasions these perpendicular splitting forces do not interact critically with the concrete tension forces (e.g. your go to example above)

I'm curious to know where we stand with that example Tom. Consider it as presented, in isolation from the retaining wall business. Is it your opinion that, in that example, the bars would still develop to yield without initiating a concrete breakout failure?

Quote (Tomfh)

Your rod-snapping and frustrum-pull-out examples do not have global tension forces aligned with the bar perpendicular splitting forces...cheating

1) As I understand it now, having bar splitting forces aligned with global tension forces, makes things worse, right?

2) So if my example doesn't have those aligned tension forces, would that not make my example more optimistic with regard to capacity?

3) If my example is more optimistic with regard to capacity, and still indicates a serious lack of capacity, is there still not a problem?

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

Consider it as presented, in isolation from the retaining wall business. Is it your opinion that, in that example, the bars would still develop to yield without initiating a concrete breakout failure?

If the breakout cones are inadequate to resist the tensile uplift then not I don't believe the bars would yield. However I suspect your case would actually yield bars, despite your loading it up with bars. You have 1000 times the area of concrete than you do steel - that equals bar snap not concrete snap. But for the sake of argument yes lets say it will fail in frustrum pullout.

Quote (KootK)

1) As I understand it now, having bar splitting forces aligned with global tension forces, makes things worse, right?

That's my understanding of it. It will break even easier.

Quote (KootK)

2) So if my example doesn't have those aligned tension forces, would that not make my example more optimistic with regard to capacity?

Yes yours is more optimistic than the bad cases and is still failing.

Quote (KootK)

3) If my example is more optimistic with regard to capacity, and still indicates a serious lack of capacity, is there still not a problem?

Your example in plain unstressed concrete. In abominable retaining wall footing we have top steel. We also have the C force at the front face of the wall strutting against the hook, making it harder to pull out.

If you believe your example is a suitable analogy, do you think adding a little bit of extra hook length is going to improve capacity?

This is where I'm at (not arguing, just trying to clarify my own understanding):

-Your examples WON'T break the bars. Cone break-out will occur if going by the numbers (although in reality the bars probably yield).
-135 hook into the bottom of an unstressed reinforced beam WILL BREAK the bar. (Wheeler)
-A 90 degree hook into the opening side (detail (a), U74 )is really bad and WON'T BREAK the bar. This is the one I would call ABOMINABLE
-A 90 degree hook into the toe (your abominable detail) is much better and may break the bar.
-We haven't completely established the difference between a 90 degree standard hook going into the toe (abominable detail #1) vs a hook with full code development length into the toe (your detail #5), That being said, Nilsson series U74, U75, U76, U70 show the range of toe lengths and hook length achieving 94%+ capacity, with marginal increase in capacity as the toe (and hook length) increases.
-Adding a diagonal bar is ideal, but a hassle for a small wall.

Quote (Tomfy)

You have 1000 times the area of concrete than you do steel - that equals bar snap not concrete snap. But for the sake of argument yes lets say it will fail in frustrum pullout.

For the sake of argument, why don't we both just stick to what we actually belive. I believe that the appendix D method that I proposed above accounts generously for the amount of concrete that surrounds the rebar.

Quote (Tomfh)

our example in plain unstressed concrete. In abominable retaining wall footing we have top steel. We also have the C force at the front face of the wall strutting against the hook, making it harder to pull out.

Agreed. There are somewhat accepted anchorage design methods to account for these things. The footing top steel bumps your phi factor up from 0.70 to 0.75. In Eligehausen's book on anchorage, he cites a method by Zhao for accounting for the influence of the compression force. I've included the method and an updated calc below. That'll get you to within about 20% of what is needed with the normal load factors included.

Quote (Tomfh)

If you believe your example is a suitable analogy, do you think adding a little bit of extra hook length is going to improve capacity?

I really do. While I know that you and CEL don't like the pulley analogy, I really think that it's the key to this. That mechanism allows the tension in the rebar to be turned (or developed depending on your perspective) without relying much on the bond stress form of anchorage. Somewhere between most and all of the rebar tension can be resisted through bearing against the monster strut coming into the rebar bend. The corner condition is kind of a unique animal. In most other STM joints, the bar force goes to zero as it crosses the restraining/developing strut. In the corner condition, the bar tension not only does not drop to zero, it may not drop at all (d_stem = d_footing).

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

For the sake of argument, why don't we both just stick to what we actually belive. I believe that the appendix D method that I proposed above accounts generously for the amount of concrete that surrounds the rebar.

I'm not too familiar with appendix D specifics sorry. I don't doubt your numbers, I'm just not sure they accurately reflect the true cone failure strength. Cone failure numbers are usually rather conservative.

Quote (KootK)

In Eligehausen's book on anchorage, he cites a method by Zhao for accounting for the influence of the compression force. I've included the method and an updated calc below.

Good to see someone quantifying this effect!

Quote (KootK)

I really do. While I know that you and CEL don't like the pulley analogy,

I was referring to your cone pullout example. Is that what you referring to?. I was asking if you thought lengthening the hooks will improve the frustrum pullout strength. I don't think it would, and I don't see how pulley analogys work there?

Quote (KootK)

Somewhere between most and all of the rebar tension can be resisted through bearing against the monster strut coming into the rebar bend.

I agree with this very much. The curve can bear onto those solid struts, as opposed to a hook turning outwards which has only unconfined concrete in tension to bear upon.

Where we seem to differ is your belief that we still need the full tension force at the exit of the bend, i.e. the pulley analogy, where we have full bar force occuring at the exit of the bend/pulley. I don't understand this at all. If there is development/bearing occuring along the curve, and most of the bar load is strutting directly off the curved portion, why this need to start all over again at the end of the curve? Why reset the development counter to zero if most of the load is already gone? It doesn't hurt to extend the bar, and personally I extend the bar all the way just for good measure. However I don't really understand why it needs it.

Consider Nilson's U76, which has a very short hook, around 200mm, and it's achieving 94% efficiency, exactly the same efficiency as U78, which has a hook around twice as long.

Just out of curiosity, what would be needed to actually test some joints?

If I made a case to my boss I might be able to donate some precast joints cast to simulate a retaining wall joint that could be tested. Heck, maybe even the University of Maine next to one of our plants might be interested and I know they have the equipment for this sort of testing. If not them, and if my boss approves, we could ship them cheap to someone willing to test them. I'd love to see this happen.

Professional Engineer (ME, NH, MA) Structural Engineer (IL)
American Concrete Industries
www.americanconcrete.com

That is an awesome idea! Sounds like a good basis of a Master's project too...

To me a full scale test is easy and simple:

- Built using small bars and thickness, say 10M and 200mm thick.
- Bars are instrumented with strin gauges before and after the bend.
- Probable cmpression strut is instrumented.
- Wall is placed and blocked in position.
- Compacted granular is placed behind the wall and an actuated UDL is placed onto the fill pushing down until we achieve failure.
- Record all data on the bars and see what is actually going on...

Quote (TME)

Just out of curiosity, what would be needed to actually test some joints?

I'd do it as a T-Joint test like Nilsson's but with the standard, inward hook detailing. In the photos below, you can see the expected crack pattern starting to form up.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (Tomfh)

I was referring to your cone pullout example. Is that what you referring to?. I was asking if you thought lengthening the hooks will improve the frustrum pullout strength. I don't think it would, and I don't see how pulley analogys work there?

No, I was referring to the retaining wall case where the pulley mechanism would come into play. I agree that the hook extensions would make little to no difference in the straight frustum pullout scenario. In my opinion, there is a significant difference between a development case and a corner joint case. With the frustum pullout being just a straight up development case, the extension would have no benefit.

Quote (Tomfh)

Where we seem to differ is your belief that we still need the full tension force at the exit of the bend, i.e. the pulley analogy, where we have full bar force occuring at the exit of the bend/pulley. I don't understand this at all. If there is development/bearing occuring along the curve, and most of the bar load is strutting directly off the curved portion, why this need to start all over again at the end of the curve? Why reset the development counter to zero if most of the load is already gone?

Yeah. For now, let's say that we're dealing with what is a "normal" retaining wall detailing configuration in my market. The stem bars turn the corner and become the toe flexural reinforcing. Here's how I see it:

1) The stem creates a moment demand (Ms) at the joint.

2) The heel contributes some of the joint moment (Mh) resisting Ms.

3) The toe contributes some of the joint moment (Mt) resisting Ms.

4) Mh + Mt = Ms

5) The existence of Ms means that there is some flexural tension in the rebar at point "A" in the diagram below. This is just M/jd stuff.

6) The existence of Mt means that there is some flexural tension in the rebar at point "B" in the diagram below. This, again is just M/jd stuff.

7) Taken together, #5 and #6 mean that there is tension on both the vertical and horizontal legs of the corner bar. It's only the imbalance that needs to be dealt with as bond stress style development. Just like the sketch that I posted before but with unequal vertical and horizontal tension in the general case.

If you disagree with any of that, please let me know which parts.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

No, I was referring to the retaining wall case where the pulley mechanism would come into play. I agree that the hook extensions would make little to no difference in the straight frustum pullout scenario.

Noted.

Quote (KootK)

If you disagree with any of that, please let me know which parts.

Agree there is flexural bending stress and hence reinforcement stress at point A and point B.

And yes if most of the moment is resisted by toe and if you using the same bar for toe and stem then OF COURSE you get the same load AsFy in both of them. It more or less resembles your pulley diagram immediately above.

However only a residual portion of the applied load at A remains in the bar at B, and vice versa. If using a hook to resist A and a straight bottom bar to resist B then you no longer need the full load at the hook tail, because the straight bar is now carrying most of that tension. I believe you are double counting in assuming you need full tension load at the exit of the pulley. The load from A doesn't get gobbled up around the hook by the struts and then reappear fully intact at the exist of the hook.

Quote (Tomfh)

OF COURSE you get the same load AsFy in both of them.

Quote (Tomfh)

I believe you are double counting in assuming you need full tension load at the exit of the pulley

The thing that I'm about to clarify may already be crystal clear to you as I've already clarified it several times above. Probably just semantics now. That said, I'm going to clarify it one more time because it's critical to this discussion:

1) WHAT I HAVE SAID: in general there will be some level of tension on both the vertical and horizontal legs of the corner bars. That level of tension may be the same but, in general, need not be.

2) WHAT I HAVE NOT SAID: the level of tension will be AsFy for both the vertical and horizontal legs of the corner bars.

Moving along... I think that we're very close to isolating the core issue now.

Please let me know what you think of these statements which I believe to be true:

1) 100% of the moment in the toe must get transferred to/from the stem.

2) #1 is true even when using standard hooked stem bars and full length footing bottom bars.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (Kootk)

wHAT I HAVE NOT SAID: the level of tension will be AsFy for both the vertical and horizontal legs of the corner bars.

But that's what your pulley diagram clearly shows? AsFy in both. Isn't that the whole reason you call a regular hook an abomination? Because you say it doesn't give full development length along the base of the footing, and thus doesn't resolve the stem load?



1)


and 2) YES

Quote (Tomfh)

But that's what your pulley diagram clearly shows? AsFy in both

But now, given all that follows, it is clear that I do not expect it to be AsFy tension at both the horizontal and vertical legs, right?

Quote (KootK 26 Sep 16 17:32)

The tension on the vertical leg could potentially take on any value from zero to As x fy. Same goes for the tension on the horizontal leg.

Quote (KootK 27 Sep 16 00:48)

Who says they're the same? Certainly not me...Because, in the general case, you're not taking it all out at the exit bar.

Quote (KootK 27 Sep 16 01:10)

I don't know what to say Tom. I've already clarified, repeatedly and at length, my opinion that 2X is the upper limit and that there is no need for symmetry.

Quote (KootK 3 Oct 16 14:46)

7) Taken together, #5 and #6 mean that there is tension on both the vertical and horizontal legs of the corner bar. It's only the imbalance that needs to be dealt with as bond stress style development.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

it is clear that I do not expect it to be AsFy tension at both the horizontal and vertical legs, right?

I thought you did expect it. Hence you saying you need the full Ld + ??? lap on the horizontal.

If in fact you do not expect AsFy tension horizontally, why then is a shorter hook an abomination? Why the need for full development length on the horizontal if you agree there's a lesser tensile force on the horizontal leg?

Quote (Tomfh)

I thought you did expect it. Hence you saying you need the full Ld + ??? lap on the horizontal.

Nope. Absolutely not. I expect there to be some significant tension coming in through the horizontal leg. And we can calculate that easily as (M_toe / jd_toe). For walls with heels and toes, it will usually be less than As x fy because:

1) M_toe = M_stem - M_heel and;
2) jd_toe >= jd stem

I believe that we're now in agreement regarding the joint mechanics when the stem bars become the footing bars. In what follows, I'll assume that we're talking about the condition where the stem bars terminate in a hook and there is a separate mat of bottom reinforcing.

Quote (Tomfh)

Why the need for full development length on the horizontal if you agree there's a lesser tensile force on the horizontal leg?

Because, fundamentally, I don't believe that it is a matter of rebar development. The bottom mat of toe reinforcing must transfer 100% of it's tension around the corner and into the stem bars. As such, I see the primary function of the horizontal legs of the corner bars as being to lap splice the footing bars with the stem bars. Technically, the bars do not need to be spliced for As x Fy. They only need to be spliced for the tension in the footing bars at the splice location. Of course, most designers are just going to splice for As x fy anyhow to keep things simple.

The splice that I'm talking about is shown in the bottom right of the sketch below (diagonal struts between green and red in plan). In forcing the rebar tension to "turn the corner" so to speak, most of the job will get done via bearing of the corner bar on the concrete strut coming into the corner. What cannot be statically resolved that way will manifest itself as some magnitude of bond stress around the bend. My sketches, and the clips from the curved bar node article above show just that.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

Technically, the bars do not need to be spliced for As x Fy. They only need to be spliced for the tension in the footing bars at the splice location.

I just want to highlight this because it seems like the key to getting both of the arguments to find a common ground that I've seen so far. Though I will fully admit I'm now having trouble following each of the arguments completely.

Professional Engineer (ME, NH, MA) Structural Engineer (IL)
American Concrete Industries
www.americanconcrete.com

Quote (KootK)

Because, fundamentally, I don't believe that it is a matter of rebar development. The bottom mat of toe reinforcing must transfer 100% of it's tension around the corner and into the stem bars. As such, I see the primary function of the horizontal legs of the corner bars as being to lap splice the footing bars with the stem bars. Technically, the bars do not need to be spliced for As x Fy. They only need to be spliced for the tension in the footing bars at the splice location. Of course, most designers are just going to splice for As x fy anyhow to keep things simple.

Ok, to clarify, you want the increased hook length not to anchor the hook, but to give the bottom bars something to grab onto?

Precisely. I suppose that, in some very low demand situations, a standard hook might suffice as a splice.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

Precisely. I suppose that, in some very low demand situations, a standard hook might suffice as a splice.

I think we should assume the bars are working hard!

You posted nilsson's image T13 and T16 above, which appear to show a hairpin detail, and a pair of L-bars detail. (it's a bit hard to see exactly)

Could you please post the reinforcing details of these, and the joints capacities. I'd be interested to see the test results of how the horizontal bar's degree of continuity with the vertical bar affects the capacity.

Here you go Tom.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Great. Thank you.

You have a point that a single unbroken bar or a full tension lap provides a stronger joint than a short lap (or no lap), hence T13 having lower capacity than T2, T12b, T16. The horizontal bar force can get into the joint more directly with a continuous bar or longer lap bar.

However the T13 hairpin doesn't perform that much worse, and it has no lap whatsoever between the horizontal bar and the hairpin! So I disagree you need to provide AsFy worth of lap between the horizontal and vertical bars. It's not a simple lap case where we have to provide Ld. Even cases T2, T12b, T16 don't necessarily loo like full tension laps.

Our case with a short hook lapping with the horizontal bars presumably does better than the hairpin with virtually no lap, but not quite as good as a single unbroken bar or a fully lapping bar.

Quote (Tomfh)

However the T13 hairpin doesn't perform that much worse, and it has no lap whatsoever between the horizontal bar and the hairpin!

I think that the explanation for some of the unexpectedly good performance one gets from poorer joint details originates from the clamping action that one gets when one has both a toe and a heel in play. Under that arrangement, you're getting an averaging effect between the capacities of a weaker joint on the toe side and a stronger joint on the heel side (analogous beam column joint below). I suspect that testing on heel-less retaining wall joints would lead to a greater disparity in the results.

Quote (Tomfh)

So I disagree you need to provide AsFy worth of lap between the horizontal and vertical bars.

I did not say AsFy worth of lap. I said however much lap is required for the M_toe/jd_toe at the splice location.

Quote (Tomfh)

Our case with a standard hook lapping with the horizontal bars presumably does better than the hairpin with virtually no lap.

I think that you're underestimating the extent to which the hairpins are just improved standard hooks. I've drawn a 10M standard hook to scale over top of a 10M hairpin below. It's almost the same lap splice dimension with the hairpin having the benefit of the extra upturn on the right to help restrain the strut coming into the corner.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

If you're going to say hairpins beat hooks/cogs I can't but help conclude you're arguing for the sake of it, as you just said you aren't concerned about the vertical bars anchorage (as I thought you were), but are in fact concerned about the horizontal bar's ability to lap onto the vertical bar so it can resolve into the joint.

You said your precise concern was to give the bottom bars something to grab onto.

How does a hairpin, which provides a much shorter lap with the horizontal bar than does a standard 90 degree hook, provide more to grab onto?

Also, is that cog you've drawn truly "to scale"? I don't think it is. This here is more like what a hairpin vs a standard cog looks like:




ANotehr version of same thing:

Quote (Tomfh)

If you're going to say hairpins beat hooks/cogs I can't but help conclude you're arguing for the sake of it

In my last post, I submitted a sketch showing why I thought that the hairpin would be as good or better than the hook. I don't think that it's fair to be accusing me of being intentionally obstinate when I continue to justify my opinions with new information.

Quote (Tomfh)

How does a hairpin, which provides a much shorter lap with the horizontal bar than does a standard 90 degree hook, provide more to grab onto?

As I showed in my to-scale sketch above, the flat part of the hook/hairpin that lap splices is nearly identical for the small bars that would typically be used for a hairpin. So, as I see it, the lap length for the two arrangements is nearly identical and the hairpin has the added benefit of an extra "hook at the end of the hook" as it were. I don't see how that could do anything but help matters with regard to splicing.

I think of it as I've shown in the sketch below. For the same lap length X, it seems intuitive to me that the versions with one or more hooks would do a better job of splicing.

Quote (Tomfh)

Also, is that cog you've drawn truly "to scale"? I don't think it is.

It is indeed to scale including: member size, bar size, bar radius. Obviously, I didn't spend hours and hours making it perfect. Probably about five minutes.

Quote (Tomfh)

This here is more like what a hairpin vs a standard cog looks like... How can a 180 degree hook provide a better lap with bottom mat that 90 degree hook?

A hairpin and a 180 hook are not the same thing and do not have the same geometry. In general, a hairpin will have a longer flat spot at the bottom.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (Tomfh)

This doesn't work at all according to pulley analogy yet in reality, is almost as efficient as unbroken bar

It's not as though I've been saying that the joint has no capacity if it's not detailed and designed according to the STM/pulley analogy. What I've been saying is:

1) Capacity will be maximized and best reflect designer determined capacity when detailing is done according to the STM/pulley scheme.

2) Other schemes will have some capacity but that capacity would be a) less than the STM/pulley detailing capacity and b) difficult for designers to estimate reliably.

I previously posted an STM model for the case where the stem bars are developed but do not turn the corner. I believe that model would be relevant to the u-bar case here. Below, I`ve superimposed that STM over the photo of the cracked u-bar specimen. As you can see, they correlate nicely with regard to strut and crack formation.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

As I showed in my to-scale sketch above

I've never seen a 90 degree bend that looks like that, and it doesn't look like any of the standard cogs in the various concrete codes. 90 degree bends have a straight extension.

Quote (KootK)

So, as I see it, the lap length for the two arrangements is nearly identical and the hairpin has the added benefit of an extra "hook at the end of the hook" as it were.

These are to scale (again, yours isn't). You are claiming a hairpin 180 degree bend provides better splice to the horizontal bar than a 90 degree bend. I'm very surprised you would argue such a thing.

Quote (KootK)

In general, a hairpin will have a longer flat spot at the bottom.

Really?? Semi-circles contain longer straight lines than do straight lines?!? That's crazy talk.

Quote (KootK)

1) Capacity will be maximized and best reflect designer determined capacity when detailing is done according to the STM/pulley scheme.

Yes, capacity is maximised with a belts and braces approach.

Quote (KootK)

2) Other schemes will have some capacity but that capacity would be a) less than the STM/pulley detailing capacity and b) difficult for designers to estimate reliably.

Such joints show about 80% capacity. To get the full 100% reliably you need to start using diagonals etc.

Quote (KootK)

As you can see, they correlate nicely with regard to strut and crack formation.

Of course. The strut and tie is a simplistic model of the concrete and steel forces, so of course they correlate. But there are areas where they don't correlate, e.g. the fact that a hairpin provides 80% capacity when it should provide next to no capacity according to STM/pulley model. You think about that mismatch between your model and the results.

Quote (Tomfh)

I've never seen a 90 degree bend that looks like that, and it doesn't look like any of the standard cogs in the various concrete codes. 90 degree bends have a straight extension.

Quote (Tomfh)

Really?? Semi-circles contain longer straight lines than do straight lines?!? That's crazy talk....You are claiming a hairpin 180 degree bend provides better splice to the horizontal bar than a 90 degree bend.

Nope. I didn't say a thing about 180 bends. I'm claiming that a hairpin would be better than a 90 bend. Everywhere that I've practiced, hairpins and 180 hooks are physically different things. Namely:

1) 180 hooks have no flat spot at the apex.

2) Hairpins will have whatever flat spot is required to fill up the space between the faces of the member, less cover.

The sketches below show the difference conceptually, by somebody else's code. The 4 x dia dimension on the right would be extended to suit the member width.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Nilson's results show 180 degree bends. I've been talking 180 degrees bends. You drew 180 degree bends (in your "to scale" sketch).

Now you're trying to sneak in trombones.

Anyway KootK, like the others, I'm done here.

You're now arguing a vertical bar with 180 degree bend provides superior "longer" lap into the bottom steel than a 90 degree hook. So we're done.

Clearly you have little interest in having an honest look at this question and getting to the bottom of it if you are making such ridiculous assertions.

Quote (Tomfh)

the fact that a hairpin provides 80% capacity when it should provide next to no capacity according to STM/pulley model.

Quote (Tomfh)

But there are areas where they don't correlate, e.g. the fact that a hairpin provides 80% capacity when it should provide next to no capacity according to STM/pulley model.

Who said that a hairpin would provide next to nothing? Not me. As I said just a few moments ago, there are lesser schemes, and they have lesser capacity. Compared to 100%, I consider 79% to be a lesser capacity.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

A vertical bar with a 180 bend does not satisfy the pulley model in the slightest.

As I said, you're now arguing the vertical bar with 180 degree bend provides superior "longer" lap into the bottom steel than does a 90 degree hook. So we're done. Not going to debate with someone who maintains absurdities like that so as to defend their position.

Quote (Tomfh)

You're now arguing a vertical bar with 180 degree bend provides superior "longer" lap into the bottom steel than a 90 degree hook. So we're done.

Again, hairpins not 180 bends. They are different products in my realm.

Quote (Tomfh)

I'm done here...Clearly you have little interest in having an honest look at this question and getting to the bottom of it.

I would argue that I've done more to get to the bottom of this particular question than pretty much anyone else on this board. I don't appreciate having my contributions diminished just because you have debate fatigue.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (Tomfh)

A vertical bar with a 180 bend does not satisfy the pulley model in the slightest.

I never said that it did. In fact, I went to great trouble to develop another, non-pulley STM model to suit the case where the stem bar is only developed. It's not like there's just one STM model available for use. We generate one (or more) STM to suit each particular detailing scheme.

To clarify, the STM below is the one that I have proposed for stem bars that end in 90 hooks or 180 hooks, not the pulley STM.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

Again, hairpins not 180 bends.

You drew a 180 bends. You referred to Nillson who showed 180 bends. I referred to 180 bends. You said these provided superior lap to bottom bars than does a 90 bend.

Quote (Tomfh)

Nilson's results show 180 degree bends.

I don't understand how you could know that. They look like hairpins to me.

Quote (Tomfh)

I've been talking 180 degrees bends.

I know it. I've been trying to get you to stop talking about them all day.

Quote (Tomfh)

You drew 180 degree bends (in your "to scale" sketch).

I'll be the judge of the intent of my own sketches and I've drawn hairpins.

Quote (Tomfh)

Now you're trying to sneak in trombones.

I'm not trying to sneak anything in Tom. I tied to show you what I mean: the difference between a hairpin and a 180 hook.






I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (Tomfh)

liar

Even though I'm not sure Tomfh will read this; I reread through most of this topic and while KootK has said some slightly antagonistic comments or defended his position in a fairly aggressive way I feel that overall KootK has attempted to maintain a balanced approach to the debate while Tomfh has been increasingly argumentative and insistent that KootK is altering his argument and/or deliberately avoiding conceding points of debate. Further, near the end, while KootK did become fairly defensive (and I see nothing wrong with this) Tomfh has resorted to increasingly accusative and pejorative language, finally culminating in petty name calling.

I'll summarize this later if you disagree with my assessment. Overall I feel KootK has been wronged by your posts here Tomfh and you owe him an apology and should probably not post in this topic again if you can't debate as a professional.

Professional Engineer (ME, NH, MA) Structural Engineer (IL)
American Concrete Industries
www.americanconcrete.com

KootK repeatedly denied drawing a 180 degree hook, claming he drew a trombone hairpins. Here is what he drew:



KootK denied Nillson drew 180 bends, but drew tromboine hairpins. Here are Nilsson's drawings.





I corrected KootK, and he again denied Nilsson drew 180 hooks, and again denied drawing 180 hooks, claiming Nilsson's hooks "look like [trombone] hairpins to me", and "I'll be the judge of what I drew, I drew [trombone] hairpins"

Quote (TME)

Overall I feel KootK has been wronged by your posts here Tomfh and you owe him an apology and should probably not post in this topic again if you can't debate as a professional.

He maintained a falsehood. I gave him several chance to change tack, but he held his line, so I called him on it. If it is unprofessional of me to point it out then so be it.

Quote (Tomfh)

KootK repeatedly denied drawing a 180 degree hook, claming he drew a trombone hairpins. Here is what he drew:

And in that same post, I introduced the sketch like this:

Quote (KootK)

I think that you're underestimating the extent to which the hairpins are just improved standard hooks. I've drawn a 10M standard hook to scale over top of a 10M hairpin below.

Quote (Tomfh)

KootK denied Nillson drew 180 bends, but drew tromboine hairpins. Here are Nilsson's drawings.

- The pictures do not state whether the bends are hairpins or 180 hook.

- You have interpreted the pictures as 180 hooks.

- I have interpreted the pictures as hairpins because they are 10M and appear to extend to the cover distance.

My interpreting a diagram differently from you doesn't make me a liar Tom. "Hairpin" may just be a term that is used differently in our respective markets. For what it's worth, the blurb below is from the Nilsson document. Obviously, you don't have the whole thing so there was no way for you to know definitively one way or another. I didn't know definitively until just now either. I just checked it out now to make sure that I wan't an accidental liar. At the very least, I think that this demonstrates that my interpretation of the diagrams was at least as valid as yours.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

KootK, I apologize for calling you a liar, I should have just stuck to the facts.


I'm sure you can appreciate how frustrating it is to hear you insist that the horizontal bars needs a full lap onto the vertical bars, and then completely dismiss the fact that a 180 hook example gets close to full capacity without any such lap, and then go off on some tangent about trombones...

Quote (TME)

I reread through most of this topic and while KootK has said some slightly antagonistic comments or defended his position in a fairly aggressive way I feel that overall KootK has attempted to maintain a balanced approach to the debate...Further, near the end, while KootK did become fairly defensive (and I see nothing wrong with this)

Thanks for defending my honor TME. Seriously, you have my genuine, non-sarcastic thanks. It gets a little cold and lonely out in agitator-land at times.

This thread has a fairly wide audience, both in terms of active participants and lurkers (I hear from them through other channels). It's also the thread where my particular style of debate has been highlighted the most. Consequently, I'd like to take the opportunity to explain my style of debate. This will be one post of, what, 250 here? I'll accept the hijacker tag.

Engineers are generally good, humble folk who do their best to avoid unnecessary confrontation. And it shows here. In my estimation, most forum members just want to dole out their practical solution to whatever practical question has been asked and move on. If differing advice crops up, they're usually happy to agree to disagree.

I'm a little different. I'm constantly, and vigorously in search of the structural "truth" to the extent that there is such a thing. Whenever I see two brilliant engineers offer differing advice without taking it to the mat in a spirited debate, I see a wasted opportunity to chase down the the truth. And lets not kid ourselves, for many things, there really is a best answer. We just have to flesh it out.

Hokie66 once referred to one of my previous sparring partners as somewhat abrasive. I like that. That's how I would like to be regarded. Sometimes I'm abrasive so that I can drive my points home. Other times, I'm abrasive in attempt to cajole my sparing partners into driving their points home. Always, the goal is to try and tease out a better understanding and, where possible, a legitimate consensus.

I get that some folks find my style of debate obnoxious. And I feel badly about that when things start to boil over, truly. On the other hand, I feel that the tenacity with which I prosecute technical debates here leads to a lot of interesting and valuable technical growth. It is, and has always been, my hope that the benefits of the latter outweigh the frustrations associated with the former. Think of me like Michael Vick. You might not want to hang with me on the weekend but you gotta admit, I do keep things marching downfield.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (Tomfh)

KootK, I apologize for calling you a liar, I should have just stuck to the facts.

Quote (Tomfh)

I'm sure you can appreciate how frustrating it is to hear you insist that the horizontal bars needs a full lap onto the vertical bars, and then completely dismiss the fact that a 180 hook example gets close to full capacity without any such lap, and then go off on some tangent about trombones...

So your sorry but I'm still a flip flopping chaser of irrelevant tangents? Gosh... thanks Tom.

I'm sure that you can appreciate how frustrating is for me to hear you repeatedly misinterpreting my statements and sketches. Nowhere have I or Nilsson been discussing 180 hooks, disparagingly or otherwise. Surely you must see that by now?

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

misinterpreting my statements and sketches. Nowhere have I or Nilsson been discussing 180 hooks

Your sketch shows regular 180 bend without flat section at bottom, not a trombone.



Nilsson shows regular 180 bends without flat section, not trombones.

Did you totally miss my 4 Oct 16 23:02 post? Nilsson and I both explicitly described our bends as hairpins rather than 180 hooks.

Personally, I don't want to be done with this thread. I think that it's interesting and important and I look forward to discussing it further with any other folks who may be late to the party. And I'll continue to make myself available for that.

That said, it seems that the discussion between you and I has ceased be fruitful. How about this Tom:

1) Make one last post on this thread if you like. Say whatever you wish. And then say no more, at least not to me, about this.

2) I'll not respond to your next post. You shall have have the last word forever more, whatever you choose those last words to be.

How about it?

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

1) Make one last post on this thread if you like. Say whatever you wish. And then say no more, at least not to me, about this.

I will sum up my understanding:

KootK believes that a 90 degree standard cog turning into the toe is “abominable” (28 Sep 16 01:35):

He has presented STM/pulley models which he argues show this to be the case. After much toing and froing, and my misundertstanding his concern as being vertical bar anchorage, we arrived at his “precise” concern, which is that a standard 90 degree cog provides insufficient lap with the footing bottom bars to transfer the bottom bar tensile load around the corner up into the joint. (4 Oct 16 01:12)

I had not fully considered his concern, but nonetheless felt the pulley analogy is misleading. So I asked for Nilsson's results, which KootK kindly offered up. Nilsson’s T13 (4 Oct 16 02:01) showed even less lap between horizontal and vertical bars than a 90 degree bend provides, and yet T13 achieves 79% efficiency. KootK presumably didn’t like these numbers, referring to them as “unexpected” (4 Oct 16 03:23).

SCALE DRAWING
To help explain away this apparently anomalous results, and preserve the pulley idea, KootK produced a “scale drawing” (4 Oct 16 03:23) of a 180 degree hook superimposed on a very incorrectly drawn 90 degree hook, to show that – despite appearances - a 180 hook in fact provides *superior* lapping to the bottom bars than does a standard 90 degree hook. Hallelulia! That’s why Nilsson’s examples work! A 180 hook touching a horizontal bar is actual a very effective lap, the pulley model is saved.

TROMBONES
Also, to bolster his argument, KootK said he wasn’t even talking about 180 hooks (despite drawing one), he in fact meant trombone hairpin (not that they satisfy Ld + ??? anyway).
And regarding Nilsson’s hairpins, KootK claims that the word “hairpin” in Nilsson’s text proves it is trombones, regardless of what Nilsson has drawn, and regardless that the actual crack pattern traces out an ordinary 180 bend.


SUMMARY
T13 has negligible lap between vertical and horizontal bars, and T12a has complete bar continuity, and yet – to KootKs surprise - they provide almost identical capacity (79% vs 82%). This contradicts KootKs primary claim that the horizontal bars need a full lap onto the vertical steel in order for the joint to function, and thus undermines the basis of his claim that a standard hook is an abominable detail.

So I get the pulley concept and why that improves the joint. Quick question regarding the hairpin/180 degree hook whatever -

@kootk - your sketch (see below) is the STM explanation for why the hairpin/180 works but not as well due to the reduced "reduced flexural depth". Is that right?

EIT
www.HowToEngineer.com

KootK; while I do agree with you (both on your rigorous debate mentality and this topic at hand); I feel that Tomfh is correct about the technicality of the "hairpin". Your drawing does appear somewhat continuous in the curve (if you intended otherwise I would look at such intent, but I'm just stating what I'm observing). Further, Nilsson does seem to be showing a continuous curve in his sketches more in line with a 180° hook. While a large enough radius curve seems fundamentally identical to a "hairpin" or "trombone" as you put it, I feel it's worth pointing out that quibbling over the minutia of what is shown is missing the key points of this debate. Finally, I've seen multiple various definitions of "hairpin" that are defined as either "trombones" or "180° hooks" or even V-shaped bars (like the ones shown for supplemental reinforcement in ACI 318 appendix D for shear) so just because Nilsson did use the term "hairpin" doesn't mean to me explicitly that they were or were not 180 degree hooks as the term has too much ambiguity to it.

All that said though, I agree with your point. So, if anyone has no objections, can we try to focus on what the differences between the reinforcements gives us and why we have some strength (79%, etc.) with one detail and more strength with others. Trying to decide if Nilsson's sketches show X or Y seems important but not the key topic we should be focusing on as it is entirely subject to opinion (in my opinion). Let's not forget the ultimate goal of this thread: "what is the proper joint detail at a retaining wall". At the most basic level; wouldn't the 79% strength level be "acceptable" as long as we knew what the reduction in strength was?

Seems like identifying the cause of this reduction in effective strength is the key; not how many engineers it takes to define a hairpin.

Professional Engineer (ME, NH, MA) Structural Engineer (IL)
American Concrete Industries
www.americanconcrete.com

I'm going to attempt to summarize what I'm seeing as the crux of each persons argument. This is GROSSLY simplified; no yelling please if I misrepresent something you said. My opinion/interpretation below:

KootK: Providing anything but my detail potentially gives you a strength of X% less than the ideal full flexural capacity of the joint. This reduction in strength isn't addressed by most retaining wall designs. Therefore I declare this joint detail verboten as it's potentially dangerous unless we can understand how much strength is lost. It's also not the most efficient as my strut-and-tie model shows so you're probably going to get best results with my detail (further recommended by CRSI, ACI joint details, etc.).

Tomfh: (To paraphrase Tomfh's own summary, which was well written). T16 has bar continuity giving almost identical capacity (79% vs 82%) to the abominable detail. Thus a standard hook is not an abominable detail, it's simply a reduction in capacity at worst and nearly as efficient. If we design for that reduction then there's nothing wrong (and this is why walls aren't falling down all over the place) and saying absolutes like "we must have full lap splices" and "we must have maximum effective joint strength" isn't the only solution.

[Ducks under desk after having painted a huge bullseye and possibly caused both parties to be forced to repeat themselves to clarify my simplifications.]

Professional Engineer (ME, NH, MA) Structural Engineer (IL)
American Concrete Industries
www.americanconcrete.com

Quote (RFreund)

@kootk - your sketch (see below) is the STM explanation for why the hairpin/180 works but not as well due to the reduced "reduced flexural depth". Is that right?

Uhh... In my heart of hearts, I want to just say yes. However, this thread has become hyper sensitive to semantics. So I'll say this instead in the interest of hyper precision:

The STM is an explanation for why a stem bar that is developed vertically within the footing depth, but is not explicitly designed to transfer tension around the corner, will work but with a reduced efficiency which, fundamentally, I believe is a result of that "reduced flexural depth" that I mentioned. In the context of the previous statement, what I mean by developed vertically is one of the following:

1) A developed vertical with no hook

2) A developed vertical with a 90 degree hook.

3) A developed vertical with a 180 degree hook.

4) Perhaps one leg of a trombone style hairpin to the extent that it would function as vertical development for the stem bars.

So, yeah. Yes.

I'll point out that "trombones" is not a KootK term. It was just included in that diagram that I pilfered off of the inter-web. The silliness of the term has been used to poke fun of me and my arguments. That said, it's kinda growing on me and folks seem to know what that means so I'm just gonna roll with it. Trombone = KootK Style Hairpin.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Oh, I thought "hairpin" was hilarious and taught it to all the guys in the precast plant. I can't wait to see what they do when I call out a "trombone" on my next project.

Professional Engineer (ME, NH, MA) Structural Engineer (IL)
American Concrete Industries
www.americanconcrete.com

TME,

what specific point of kootks is it that you are agreeing with?

As for the cause of the strength reduction, as Nillson captions - hairpin anchorage failure. With continuous anchor bar or long laps you can milk a bit more capacity out of the joint.

Quote (TME)

Your drawing does appear somewhat continuous in the curve

My green rebar 90 has a definite curve in the corner. My green rebar 90 has a definite straight, horizontal extension. These things are congruous with reality. Where I went wrong is that I measured the end of the hook as 12db from the vertical bar centerline rather from the point of tangency on the curve. So yeah, the straight part is a little short (~60 mm). Hari Kari for me.

Quote (TME)

Further, Nilsson does seem to be showing a continuous curve in his sketches more in line with a 180° hook. While a large enough radius curve seems fundamentally identical to a "hairpin" or "trombone" as you put it, I feel it's worth pointing out that quibbling over the minutia of what is shown is missing the key points of this debate.

It's a small sketch. Maybe it's not drawn in incredibly intricate detail. Maybe it's incorrect of me to interpret the width of the hairpin extending to cover as meaning trombone style. Maybe when Nilsson says hairpin, he actually means 180. Maybe it is a big radius rather than a true "trombone".

- Nilsson said it was a hairpin.
- It looks like a hairpin to me.
- There is nothing that explicitly said that it is a 180.

While my interpretation may indeed not be correct, when you add all of that up, it leads me to believe that my interpretation is at least as valid as anyone else's. Additionally, when I posted my apparently ambiguous sketch, I included text clarify my intent. And then, subsequently, I re-clarified it over and over again. I don't see how, at this point, anyone could be unclear with regard to what my intention was.

Quote (TME)

I feel it's worth pointing out that quibbling over the minutia of what is shown is missing the key points of this debate.

I agree but I can only respond to the points that people choose to query me on.

Quote (KootK)

At the most basic level; wouldn't the 79% strength level be "acceptable" as long as we knew what the reduction in strength was?

Yup. And I have previously mentioned that sometimes increase the rebar by 1/%Efficiency rather than include the diagonal bars for small retaining walls where I'm unconcerned with crack width. One point of concern for me, however, is how reliable the inefficiencies scale up for larger bars. All of the testing was done on 10M. And my MathCAD work above suggests that things may get worse, quickly, for larger bars.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (TME)

KootK: Providing anything but my detail potentially gives you a strength of X% less than the ideal full flexural capacity of the joint. This reduction in strength isn't addressed by most retaining wall designs. Therefore I declare this joint detail verboten as it's potentially dangerous unless we can understand how much strength is lost. It's also not the most efficient as my strut-and-tie model shows so you're probably going to get best results with my detail (further recommended by CRSI, ACI joint details, etc.).

Firstly, what the hell are you doing man?!? I just about had this settled down.

Secondly, yeah, you pretty much nailed it. Minor clarifications:

1) I might have toned down some of the absolutes for mass consumption.

2) I believe the 90 degree hook termination to be a fundamentally poor detailing choice, regardless of the efficiency numbers etc. Anecdotally, every time that I've encountered a proponent of that detail, that proponent has proven themselves to harbor misunderstandings about how concrete joints function.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (Tomfh)

Your "scale drawing" went wrong in at least three ways- The bend diameter is all wrong. The straight extension is all wrong (you've drawn 4db not 12db). The overall length is all wrong (a standard 10M 90 degree bend within a 200 wall with 25 cover will extend past the wall face).

I thought that we had a truce in place Tom? Can we not abide by that? Self discipline, it's what separates us from the animals.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (Tomfh)

Yes it's fairly clear by now that your beliefs are impervious to the results.

I'm just going to respond to others while doing my best to stick to the truce Tom. Please don't interpret that as any kind of slight as that's certainly not my intent.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

Therefore I declare this joint detail verboten as it's potentially dangerous unless we can understand how much strength is lost.

Well then you're going to have to ban just about all the joints in this table, including the case of an unbroken L-bar, as the only joints which reliably give you 100%+ are the ones with diagonals and ties bandaging the knee up.:




The 79% is in the ball park for all these sort of joints.

Quote (Tomfh)

Quote (KootK)
Therefore I declare this joint detail verboten as it's potentially dangerous unless we can understand how much strength is lost.

I just want to point out that this was my interpretation of KootK's position. While he agreed with my assessment, I want to make sure people don't think this is a direct quote of KootK.

Professional Engineer (ME, NH, MA) Structural Engineer (IL)
American Concrete Industries
www.americanconcrete.com

Quote (Tomfh)

TME, what specific point of kootks is it that you are agreeing with?

First, let me point out that this is half my opinion and half my understanding/conclusions based on weighing all the comments presented here. I will fully admit I am no expert in concrete joint detail though I believe I have enough of an understanding to draw my own conclusions. Also, while I share KootK's abrasive debate style I do not have any "favorite" person here and considered all positions equally.

This is my own interpretations of the arguments presented, please correct my if I misunderstood someone's position. That said, these are the points I believe KootK and I agree on.

• Failure of a straight or hooked bar which is fully developed does not preclude an "appendix D" style tension pull-out cone. While it may not control the strength of the joint, not checking the capacity of the frustum pull-out failure is neglecting a failure mode. Providing force transfer around a corner is one way to ensure that this failure mode will not control. It's worth pointing out that I believe even a 90° hook or straight dowel will still transfer some force around a corner, as both KootK and pretty much everyone else has indicated.
• Curved bar nodes provide for force transfer around corner joints. Force transfer around the corner is required for the cantilevered retaining wall to function as idealized.
• Fully developing a large radius curved bar around the joint as KootK shows appears the be the most efficient detail when comparing tested results, modeling of forces, recommendations from CRSI/ACI/etc., constructibility, crack control, and ease of analysis. This is probably the biggest reason I'd side with KootK's argument over others.

Things I may not agree on fully with KootK's argument.

• There are other ways to get strength out of the joint. A simple 90° hooked bar or a hairpin will transmit some force around the corner and likely will work sufficiently. However, considering the full flexural capacity of the joint without taking in account the potential deleterious effects of this setup is not appropriate. Use of this detail, properly accounting for all the strengths, is perfectly acceptable. I do agree with KootK that confirming that you have properly scaled Nilsson's test results is difficult without understanding the reason for the loss in strength.
• I'm not convinced that CELinOttowa is correct in his force diagram for the curved node bar forces where he included the development forces. But, neither am I convinced he's wrong. I believe there's some definite merit to his comments regarding the forces involved.

By all means discuss some of the above with me; but I don't want to derail this thread by forcing people to stop and repeat what they said to correct my interpretation.

Professional Engineer (ME, NH, MA) Structural Engineer (IL)
American Concrete Industries
www.americanconcrete.com

Quote (TME)

There are other ways to get strength out of the joint. A simple 90° hooked bar or a hairpin will transmit some force around the corner and likely will work sufficiently. However, considering the full flexural capacity of the joint without taking in account the potential deleterious effects of this setup is not appropriate.

So I woke up this morning, popped out of bed, and realized that I still have a bit more to give on this. And not just petty bickering either. Good stuff.

Everybody seems to want to know how, and how well, the joints work when the stem bars are terminated in a developed hook and no more. Above, I've pitched two explanations for the mechanism/efficiency in such joints: the STM model and the joint clamping mechanism analogous to a perimeter beam column joint.



The major drawback with STM models, of course, is that nobody has time for that stuff in a design office. Additionally, there are some features of the joint behavior that are actually less intuitive when presented in STM than they might otherwise be if presented in terms of traditional sectional design methods.

So... I've decided to generate a proposed sectional method for the design of these joints as illustrated in the two sketches at the bottom of this post as follows:

1) Detail A shows the excellent clamping mechanism that could be relied on if only the strut would not slip off of the stem bars due to lack of development. I've assumed that the tension forces in the footing rebar can be conceived of as being pulled across the joint and manifested as additional compression on the other side. We do similar things in strut and tie so I'm comfortable with the validity of the assumption. Really, it's a big part of what makes the clamping mechanism so effective.

2) Detail B shows a proposed method of adjusting the design for the amount of stem bar development actually available. Again, I've assumed that the tension forces in the footing rebar can be conceived of as being pulled across the joint and manifested as additional compression on the other side. Basically, it amounts to this:

2a) Figure out the dimension La to make things pan out. Some iteration might be required.

2b) Increase the heel reinforcing to account for the reduced flexural depth (jd_heel_reduced).

2c) Increase the stem reinforcing to account for the partial rebar development (La vs Ldh).

2c) Check one way shear on heel based on reduced flexural depth.

2d) Ignore the compression strut as we would with other sectional methods (I guess).

2e) Have faith that the system will "find" this load path rather than initiate an anchorage failure along the way.


An interesting feature of this is that it provides a couple more ways to explain why we don't see failures with this detailing:

1) We know that current, code specified development lengths are quite conservative. The reduced flexural depth in the heel is probably better than it seems. And the ratio La/Ldh is probably greater than it seems.

2) We know that the probable yield strength of rebar closer 1.25 x fy. This will help to offset the additional rebar required by the inefficiencies inherent in the joint.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (TME)

Fully developing a large radius curved bar around the joint as KootK shows appears the be the most efficient detail when comparing tested results, modeling of forces, recommendations from CRSI/ACI/etc., constructibility, crack control, and ease of analysis. This is probably the biggest reason I'd side with KootK's argument over others.

I agree an unbroken bar is more more efficient. It is slightly more efficient. The hairpin (T13) gives 79%. When all else is equal and the bar continues into a decent length lap (T12b) the capacity goes to 82%. Different reinforcement ratios (T12, T16) can see the lapped/continuous bar capacity increase up to 100% capacity (perhaps the hairpin case would increase too, or maybe it maxxes out around 80%)

What I disagree with is that this proves the STM/pulley model, that it proves a bar with less than full lap is "abominable" and that it proves simply wrapping a continuous bar "guarantees" idealized performance. It is a silly to say 79% reflects a dangerous joint and that 82% is perfectly good reliable joint that we needn't look twice at. If one is questionable, the other is too.

Quote (TME)

There are other ways to get strength out of the joint. A simple 90° hooked bar or a hairpin will transmit some force around the corner and likely will work sufficiently.

Yes I agree a standard hook will provide some lap strength and would help keep the joint tied up. Such a case lies midway between the 180 hairpin vertical bars with no lap and a continuous 90 degree bar, so you expect intermediate performance.

Quote (TME)

However, considering the full flexural capacity of the joint without taking in account the potential deleterious effects of this setup is not appropriate. Use of this detail, properly accounting for all the strengths, is perfectly acceptable. I do agree with KootK that confirming that you have properly scaled Nilsson's test results is difficult without understanding the reason for the loss in strength.

I'm all for testing and understanding joints. It just strikes me as a real double standard to condemn less than full lap as dangerous whilst maintaing that a continuous bar is safe and reliable. How can 79% be dangerous and 82% safe?

As for the reason for the loss in strength in the hairpin - it is anchorage failure, as noted on nilssons images. The hairpin can pull out easier than a continuous/90degrees bar beacause there is nothing crossing the pullout interface - see image below. This is in line with KootKs ideas, but is not nearly as bad as he imagines.

Since the continuous bar cases start failing at only 3% more load, at 82%, these joint has other problems anyway. You can't simply make a continuous bar, dust your hands, and give yourself a pat on the back.

If you want the guaranteed 100% performance you are demanding, with no questions as to reduced capacity, you need diagonals and bandage bars, etc

Fixed.

Quote (Tomfh)

Different reinforcement ratios (T12, T16) can see the lapped/continuous bar capacity increase up to 100% capacity (perhaps the hairpin case would increase too, or maybe it maxxes out around 80%)

This was my thought; if demonstrated otherwise I would happily concede that point.

Quote (Tomfh)

What I disagree with is that this proves the STM/pulley model, that it proves a bar with less than full lap is "abominable"

Agreed on both.

Quote (Tomfh)

and that it proves simply wrapping a continuous bar "guarantees" idealized performance. It is a silly to say 79% reflects a dangerous joint and that 82% is perfectly good reliable joint that we needn't look twice at. If one is questionable, the other is too.

Agreed to some degree. While the highest value of the tested "hairpin models" and the lowest values of the tested "KootK Hook™" do almost coincide; the impression I got was improved performance was achieved with the KootK Hook™. But, yes, establishing why there was overlap in the tested values is clearly important and doesn't mean that it's an infallible joint detail. What I meant by "guarantees" idealized performance was that it matches a model that makes sense to me (not saying others are wrong, just that I understand KootK's pulley model best), and it provides rebar continuous around the joint avoiding plain concrete areas that can be broken out in a frustrum "appendix D" style failure mode, something I feel to be quite beneficial. I clearly worded that poorly as I was not trying to imply that it was a perfect joint detail that by bending the bar towards the toe and developing it you automagically achieve 100%+ efficiency.

Quote (Tomfh)

How can 79% be dangerous and 82% safe?

Agreed, it can't. However, I feel that if I took nothing else from this thread; knowing that the KootK Hook™ detail provided 3% or more improved performance in a joint with no downsides then it's at least better than nothing. But, as KootK said, I'd wish to find the truth of what is actually happening in the joint and why both methods have less than 100% efficiency in some tests, what's the maximum possible efficiency out of each style, and what needs to be done to ensure 100%+ all the time.

Quote (Tomfh)

As for the reason for the loss in strength in the hairpin - it is anchorage failure, as noted on nilssons images. The hairpin can pull out easier than a continuous/90degrees bar beacause there is nothing crossing the pullout interface - see image below. This is in line with KootKs ideas, but is not nearly as bad as he imagines.

Agreed on both points. Failure matches what I feared/expected, but neither was perfect.

Quote (Tomfh)

If you want the guaranteed 100% performance you are demanding, with no questions as to reduced capacity, you need diagonals and bandage bars, etc

Perhaps this is the answer. But, yes, if these joints have been a concern in my designs this has been my approach.

Professional Engineer (ME, NH, MA) Structural Engineer (IL)
American Concrete Industries
www.americanconcrete.com

Hi All,

Interesting thread!

My understanding is these "tests" and all the research has been completed for quite a thin (8") wall joint?

What happens with these details if we are talking about a much larger concrete section, say in the stem of a larger wall or a large industrial concrete tank? These section may increase to as much as 18-20" thickness, in these cases are we still looking at the 60-80% joint efficiency with the same or similar details? Or can we increase the joint efficiency without adding the diagonals just because the hooked or hairpin bars for sure are developed past their intersection point in the case of an opening joint?

Sweet, some questions that are easy to answer. Well... for the most part.

Quote (Everynameistaken)

My understanding is these "tests" and all the research has been completed for quite a thin (8") wall joint?

- For the retaining walls, it was 8" walls and 10" footings.

- For the T-joints is was an 8" x 8" post hanging down from a beam member of unspecified depth.

Quote (Everynameistaken)

Or can we increase the joint efficiency without adding the diagonals just because the hooked or hairpin bars for sure are developed past their intersection point in the case of

Are you inquiring about the horizontal joint between the walls and the foundations? Or there vertical corner joint where two walls meet?

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Hi,

I think both, is there really that much different between the two except perhaps where we put our construction joint? If they are the same thickness then very similar.

Both flexural members with an opening type joint applying tension to the inside face bars, yes the base to wall stem also could have some compression I guess?

Quote (Everynameistaken)

I think both, is there really that much different between the two except perhaps where we put our construction joint? If they are the same thickness then very similar.

There's a difference. The retaining wall joints benefit from having both a toe and a stem. You get that helpful clamping action that I pitched a few posts back. With the corners, particularly the opening corners, clamping goes out the window and the importance of the rebar detailing becomes exacerbated. The Nilsson data for corner joints is shown below.

The question of scale is an interesting one. As I mentioned previously, where the detail would rely on concrete in tension (breakout), the mathCAD work that I posted above indicated that that would get worse with larger bars. So take that for whatever you think it's worth.

With regard to the size of the concrete members, I'm torn between which of these statements would be most applicable:

1) Member size should scale up in proportion to the rebar tension force (db^2, spacing) in order to keep the efficiency the same. If your members are doubling in size and you're still using small bars at a reasonable spacing, I suspect that your inefficiencies would improve.

2) Member size should scale up in proportion to the rebar diameter (db^1) in order to maintain the same efficiency. As bars get bigger, the bend radii get bigger too and you start pinching your joints if the joint dimensions aren't also scaling up proportionally.

I lean more towards #1 being the dominant effect. That's just my personal opinion though. I've no hard data to back it up.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Johansson did some work in 2001 that is interesting as well. Not sure why Nordic folks seem to rock at this. Johansson's studies focused more on closing joints than most other researchers too which is interesting. Basically, as long as you've got a well spliced corner bar set on the outside, the closing joints perform great.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Oh, we're going to need a whole entire new thread if you start talking about closing joints KootK. :P

Professional Engineer (ME, NH, MA) Structural Engineer (IL)
American Concrete Industries
www.americanconcrete.com

Quote (TME)

Oh, we're going to need a whole entire new thread if you start talking about closing joints KootK.

Ahh... but we've been been talking about closing joints the whole time TME. That's the "pulley". Might as well go for an even 300 here I say!

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

It is and it isn't a closing joint.

Eg retaining wall U76 is far more opening than closing.

I should have said "closing corner joints", and it was meant mostly in jest. Obviously there are important parallels, though as Tomfh said it's a combination of opening/closing (tee joint) so parallels but not directly comparison seem appropriate, agreed?

Professional Engineer (ME, NH, MA) Structural Engineer (IL)
American Concrete Industries
www.americanconcrete.com

Quote (TME)

it's a combination of opening/closing (tee joint) so parallels but not directly comparison seem appropriate, agreed?

I agree that every retaining wall with a stem and a toe is some part opening joint and some part closing joint. That was really my point. Surely you'll give it to me on credit that I did not think that a retaining wall was just a simple closing joint?

Quote (TME)

so parallels but not directly comparison seem appropriate, agreed?

Not agreed. The toe to stem part of the joint absolutely is a closing joint and visualizing it as such is the key to recognizing the appropriate detailing.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Some of these walls can contain closing joint, but not always.

U73, U74, U76 opening only.

U70, U75, U77, U78 opening and closing

Quote (KootK)

Surely you'll give it to me on credit that I did not think that a retaining wall was just a simple closing joint?

Of course.

Quote (KootK)

Not agreed. The toe to stem part of the joint absolutely is a closing joint and visualizing it as such is the key to recognizing the appropriate detailing.

While I think I see what you're saying; this quote above appears contrary to your other quote at the beginning of this post. Can you clarify?

Professional Engineer (ME, NH, MA) Structural Engineer (IL)
American Concrete Industries
www.americanconcrete.com

Quote (TME)

Can you clarify?

I'll certainly try. I see every conventional retaining wall joint as a superposition of two joints: a toe-stem closing joint and a heel-stem opening joint. And we know coming in just how much moment transfer capacity is required of each.

Whatever moment resides in the toe has to be transferred around the closing joint and into the stem. If one wishes to stick with RC concrete principles rather than relying on concrete in tension, as I do, then the flexural tension in the toe rebar needs to round the corner into the stem via a capable corner bar and what we've been calling the pulley STM (my stuff and Johansson's above).

So at the end of the day, the portion of the retaining wall joint that functions as a closing joint is governed by the same mechanics and rational detailing as any other (non-T) closing joint. Thus I see the two as pretty much perfect analogs.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

If we want to arrive at the truth of how concrete works, it's worth acknowledging that all reinforced concrete - the pulley STM model included - relies fundamentally on concrete in tension.

Bar anchorage, shear, "phantom ties", the very fact that concrete is solid - all of this relies fundamentally on tensile forces. When we reinforce concrete we aren't eliminating tension, we are spreading and redistributing the tension out within the concrete so that it is manageable and hopefully the steel is the critical element. But we're still relying on it.

Take the pulley STM model - look at how it fails - tensile overload in the diagonal direction.





These cases start failing at 82% due to concrete tensile overload - failure of the so called "phantom ties". (some give 100% too of course, because the concrete tensile stresses haven't been exhausted).

To get the full 100% guaranteed performance you need additional transverse ties to clamp those cracks.



However those transverse tie too rely on their own anchorage - more tensile stress in the concrete!, and so on. By that stage you've gotten to the point you can start ignoring it and pretend you aren't relying on concrete tensile forces.

But if we want arrive at the truth of how it all this stuff work, we need to let go of the comforting belief that we aren't relying on tension in the concrete. In reality, tension keeps it standing.

This might be somewhat relevant - I was reading through Nawy's reinforced concrete book on Corbel deign via STM approach and he states that "the top bars in one layer have to be fully developed along the longitudinal column reinforcement. This would seem to go along with the pulley concept, no?

Still struggling with the best way to upload/markup create and share image... so see attached.

EIT
www.HowToEngineer.com

• http://files.engineering.com/getfile.aspx?folder=ee30c7e5-3faf-46fd-90cb-fe

Quote (RF)

This would seem to go along with the pulley concept, no?

Bingo. Nice add. And you'll see folks using standard hooks in that application too.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

1) Seems to me there should be a good sized anchor bar inside the bend of the primary tension steel at Point A to spread out the 113,000# compression of strut AC.
2) The #3 anchor bar seems much too small to develop the primary tension steel at the outside of the corbel.
3) There is no mention of anchorage of the 2-#6 framing bars at point D. Do they just stop at the column steel?
4) Compression strut DC' ends up with no tie to resist the horizontal force at C'. Perhaps you need tie D'C' for that purpose.

BA

@RFreund: you'd asked about exterior beam column joints before. I stumbled upon this and though it might interest you: Link

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Great report.

Seems be saying that bar stress at the start of the tail remains very low until the joint starts spalling and falling apart, at which case the tail start working hard and presumably keeps it all hanging together a bit longer. Hence nilssons getting 80-100%+ from pulley reinforcement rather than 79% with 180 degree hook (discontinuous pulley).

Quote:

If we want to arrive at the truth of how concrete works, it's worth acknowledging that all reinforced concrete - the pulley STM model included - relies fundamentally on concrete in tension.

Bar anchorage, shear, "phantom ties", the very fact that concrete is solid - all of this relies fundamentally on tensile forces. When we reinforce concrete we aren't eliminating tension, we are spreading and redistributing the tension out within the concrete so that it is manageable and hopefully the steel is the critical element. But we're still relying on it.

Good point, and I think it's something we all tend to forget.

In another context, I have been looking at research papers on tension stiffening effects, and it's quite interesting how virtually all of them (even the most recent ones) describe the strain effects of micro-cracking around the reinforcement, then go right back to talking about "bond slip" as thought the concrete somehow slips over the surface of the steel, in spite of the ribs.

Doug Jenkins
Interactive Design Services
http://newtonexcelbach.wordpress.com/

Quote (IDS)

micro-cracking around the reinforcement, then go right back to talking about "bond slip"

One of the first things you learn in uni is that concrete is weak in tension, and that GOOD engineers never rely concrete in tension. So most engineers are quite good at deflecting our reliance on concrete tension without missing a beat.

Another interesting hint at this discussion occurs in the ACI318-08 Commentary R12.10.6. When discussing a corbel they say that:
"An end hook in the vertical plane, with the minimum bend diameter bend, is not totally effective because an essentially plain concrete corner will exit near loads applied close to the corner. For wide brackets perpendicular to the plane of the figure (again the figure is a corbel) and loads not applied close to the corners, U-shaped bars in a horizontal plane provide effective end hooks."
Not sure how that eliminates the plain concrete corner though.

EIT
www.HowToEngineer.com

@RF: I believe that those statements are in reference to the end of the corbel farthest from the column.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Alright, admit it: Who else got crazy busy with the "get this done this year!" rush and forgot about this?