top plate on steel beam 1

[JStructsteel]

When replacing a load bearing wall with a steel beam, how thick of a top plate on the beam do you use? I have used just a single 2x before, but had a few contractors ask if they can do a double top plate.

For a roof, I get it to nail any uplift brackets. For a floor, any advantage other than fastening?

What do you folks spec?

Usually I thru bolt too with 1/2" bolts on a staggered pattern, 18" o.c.

 

[SE2607]

We usually do not use a nailer and just nest the 2x10's into the web and block between them (assuming joists coming from both sides of reasonably equal load). Saves a bunch as no fabrication is required.

There is the same issue with bolt/joist interference when bolting the packing to the steel beam.

 

[SE2607]

Meh, I doubt anyone actually looks at this.

Speak for yourself, please.

Deflection controls 99% of beams for me anyway and a few holes ain't going to affect that.

I agree that bolt holes won't have a significant impact on deflection.

Most of my customers use steel beams because they are shallower than an equivalent LVL. Adding 2" extra of nailer (4x versus 2x) would be a deal breaker.

I believe I alluded to my avoidance to 4x nailers for similar reasons.

Sometimes on dropped beams I don't even use a nailer -

I believe I alluded to the same practice. mostly because a 2x nailer prevents a W8 fitting with 2x10 joists and not being able to get a 2x nailer to work because there isn't enough room for a 1-1/2" nail.

just have the holes drilled under the joists and lag directly into them.

Not sure I understand this, but it's obvious we have different practices.

 

[SE2607]

Meh, I doubt anyone actually looks at this.

Those who don't should if strength governs. A 9/16" hole in a 4" flange represents 14% loss of the flange area. OTOH, I have a colleague and a long time friend who I accuse of over engineering so much, he doesn't need to do calculations.

 

[XR250]

Those who don't should if strength governs. A 9/16" hole in a 4" flange represents 14% loss of the flange area. OTOH, I have a colleague and a long time friend who I accuse of over engineering so much, he doesn't need to do calculations.

How many situations have you actually run into where a steel beam WITH a wood nailer was controlled by strength? I suppose if you are designing to code minimum deflections, than, yea it could come up. So maybe I am like your colleague?

"

just have the holes drilled under the joists and lag directly into them.

"Not sure I understand this, but it's obvious we have different practices."

Push the bare steel beam tight against the joists and run lags up into the bottom of the joist thru the top flange. Requires a bit more planning as the holes have to be located accurately, but saves 1 1/2" when needed.

 

[dianium500]

I use both single and double plates. I like the double for roof connectors because it allows for the use of hurricane connectors like H10s. Also, when hanging joist with top flange connectors, it's better to have that extra plate in my opinion.

 

[SE2607]

How many situations have you actually run into where a steel beam WITH a wood nailer was controlled by strength? I suppose if you are designing to code minimum deflections, than, yea it could come up.

Not many, but, as I wrote previously, it is a business decision on my part, one that not many admit to on this forum but would rather rationalize as "professional judgement". I don't want my clients, mostly contractors, to think they can drill holes through beam flanges. If and when I come into a condition where the strength is compromised and I won't permit holes, I can hear them (in my mind) saying "you let us drill holes on xxx project, why can't we do it on this one?".

So maybe I am like your colleague?

I meant that as tongue and cheek as I do with my colleague. I certainly don't mean to offend how you practice engineering. I do admit, I try to squeeze every nickel out of project. I once complained on this forum that I got a plan check correction for being 1% over stressed and got hammered for designing so close. When I told my client about this, he reported to the home owner of a project we both worked on and told them "See? Bill doesn't over engineer!". So, most of my customers appreciate my approach and respect my "business decisions".

"

Not sure I understand this, but it's obvious we have different practices.

Push the bare steel beam tight against the joists and run lags up into the bottom of the joist thru the top flange. Requires a bit more planning as the holes have to be located accurately, but saves 1 1/2" when needed.

That's pretty slick. Never thought of that before. Would you do that with a SDS screw (1/4" diameter)?

 

[XR250]

That's pretty slick. Never thought of that before. Would you do that with a SDS screw (1/4" diameter)?

Yes, that is my go-to screw for steel to wood fastening.

I meant that as tongue and cheek as I do with my colleague. I certainly don't mean to offend how you practice engineering

No offense taken. I admit, and others have pointed out, I am somewhat of a cowboy.

 

[sdz]

It might depend on why the builder wants to use a double timber plate instead of a thicker single plate. Why not a 3” plate?

If a double plate there could be a problem with uplift if the superstructure is connected to the top plate and the beam is connected to the lower plate. You need to make sure there is effective load transfer.

If shrinkage is a concern use seasoned timber.

Instead of using bolts consider using TEKS® 5 self drilling screws. They’re faster and easier to install, even if you have to predrill. Countersink if necessary.
e.g. TEKS® 5 HWH CL 1/4-28 X 4" screw will fasten a 3” timber plate to a ½” flange.

 

[Eng16080]

Typically, my detail has a single 2x wood nailer with 9/16” flange holes staggered at 12” o.c. (one bolt every 12”). The nailer is connected with 1/2” bolts with washers and a nut.

I prefer through bolts over screws, PAFs, and other fasteners. I’ve even had contractors request bolted connections because they’ve had problems installing these other types of fasteners.

I don’t normally consider the section loss due to bolt holes. Since the holes are sparsely spaced, I don’t think the true strength would be equivalent to a top flange with a continuous 9/16” strip missing. I suspect it would be much higher. Also, the wood nailer would likely increase the overall bending capacity, offsetting some of the loss due to the holes. I don’t have an in depth analysis of this, but my judgement tells me the effect due to holes can be neglected in this case.

Also, prior to a bending failure, I think the beam would likely deflect to the point that the small gaps between 1/2” bolts and 9/16” holes would close, therefore restoring the net section loss.

 

[XR250]

Also, prior to a bending failure, I think the beam would likely deflect to the point that the small gaps between 1/2” bolts and 9/16” holes would close, therefore restoring the net section loss.

I suppose we can start specifying 1/2"Ø holes for 1/2"Ø bolts. Not sure why we go up 1/16" in this case. I have never had an issue of them fitting in my own fabrications.

Also, this...

Section E3.6g.5 of the 2010 Seismic Provisions for Structural Steel Buildings (ANSI/AISC 341-10) allows bolted column splices in special moment frames. It states: “Bolted column splices shall have a required flexural strength that is at least equal to Ry Fy Zx /αs of the smaller column, where Zx is the plastic section modulus about the x-axis.” However, it seems that any holes placed in the section will reduce its strength below the expected strength. Increasing the size of the column or reinforcing it with plates is not helpful since Zx will be increased, resulting in an even higher demand. How can a bolted splice be used in a special moment frame?

Your question contains an incorrect assumption. It has been shown through physical tests that within certain limits, a member with holes can still develop its gross flexural strength. Section F13.1 of the Specification for Structural Steel Buildings (ANSI/AISC 360, available at www.aisc.org/specifications) addresses “Strength Reductions for Members with Holes in the Tension Flange.” In this section, when Fu Afn ≥ Yt Fy Afg the limit state of tensile rupture does not apply. For capacity-based design, this relationship would have to be adjusted to account for Ry and Rt . For ASTM A992, both Ry and Rt equal 1.1, so the relationship remains unchanged. AISC’s Prequalified Connections for Special and Intermediate Steel Moment Frames for Seismic Applications (ANSI/AISC 358-10 with ANSI/AISC 358s1-11, available at www.aisc.org/ seismic) also addresses bolted flange plate connections, which will force yielding in the beam even though there are holes in the beam flanges. There are several factors that might help explain this observed behavior. The tensile strength is greater than the yield strength; this offsets the presence of the holes to some degree. The holes on the compression side have little impact on the strength, and the maximum compression strength, though commonly assumed to be equal to the yield strength, is likely greater than we assume. Larry S. Muir, PE