Reinforced Concrete Oneway Slab with Pattern Loading

[hdn32]

Hi,

I am checking an existing elevated floor slab being used as machine shop for the installation of a CNC machine. The building was built in the 1950s.

According to the record drawings, the elevated (4-span) slab was designed for 200psf superimposed live load.
My calculation and model confirm that the slab is capable of supporting 200psf only when loaded in ALL 4 spans and does not have enough capacity for 40psf of superimposed LL when applied pattern loading. The moment envelop is way outside the rebar moment capacity in tension zones above the 3 interior supports (transverse beams).

Have anyone run into this same situation? Does the result make any sense at all?

Please share your experience regarding what could be done to the floor slab/building.

Thank you,

hdn

 

[JAE]

Looking at AISC's four span diagrams (Table 3-23 in the 13th edition)

First interior support moment
- Continuous loading M = .107wL^2
- Adjacent loading M = .121wL^2 --- a 13% increase in moment

So I'm not sure how a 40 psf live load alternated can exceed a continuous pattern moment based on 200 psf.

There is solid dead load on the span (continuous) so the pattern effect on total moment is even less.





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[EngineeringEric]

If modeling this. Did you perhaps add an additional 40 psf on top the design 200psf? I would also think that the older 1950 design would be more ASD which typically has more steel than LRFD does currently (so i have been told)... Don't quote/flog/design on this, it is just something i was informed from some senior bridge guys who would check and reduce bridge loading capacity.

 

[jike]

I am surprised. Why are you using 40 psf pattern loading?

Have you taken rebar requirements at the face of supports?

 

[hdn32]

Thank you for the comments/questions.

@JAE: Yes, I totally agree with that analysis part. However, the problem is the lengths of the top reinforcement. With a certain pattern/(skip live load), the width of tension zone (due to negative moment) change and it will encroach portion of the slab that does not have top reinforcement or the top reinforcement has not developed the full development length yet.

I have the same problem with another area of floor, where it was designed for 500psf of superimposed LL and the slab is 8in thick with bigger rebars (but same length with the 200psf floor which confirmed by a scan survey).

As far as the 500psf floor, there is a suggestion from CRSI manual (based on ACI 318-89) page 7-1, which read "The designer should, for example, extend some top bars to be continuous throughout the span if ratio of live to dead load materially exceeds three"

@Engineering Eric: No, I did not apply 40psf on top of the 200psf

@Jike: I use pattern load, because it is the nature of live load

All: I did contact PCA/Structural Point to have their engineers review my assumptions, models and hand calc. After more than two hours on the phone reviewing drawings and models line by line, we have arrived with same conclusions.

Since this is going a "big call", I really want to cover all the aspect and find out if there is anything I can do to this floor

 

[KootK]

How much superimposed load capacity do you need? A good strategy might be redistributing your negative moments. Given the large original design loads, you should have considerable capacity available even if you treat the spans as simply supported. I've reviewed more than one old building where the top steel was just nominal crack control reinforcing.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[hdn32]

KootK,

Thank you for your response.

I would need 100+psf since this is a machine shop. However, there are heavy CNC machines (20000 lbs - 40000 lbs) installed on the floor mostly above floor beams.

Could you clarify your comment "redistributing your negative moments" for me please? I am familiar with ACI's moment redistribution and is this what you meant?

As far as considering as simply-supported span: yes. I have been on it, but the slab was cast integrally with transverse beams so there is no rotational displacement allowed and the detailing of the existing slab/beams were never intended for pinned end conditions. Therefore, I am worried that with flexural cracks at face of support/transverse beams due to negative moments, the concrete section area will reduce and it may fail in shear (brittle failure).

Could you please point me to any reference/direction as to figure out "top steel was just normal crack control reinforcing"?

Regards,

hdn

 

[JAE]

hdn32 - of the many older concrete structures I've analyzed over the years the rebar development always rears its ugly head.
The older codes (and former engineers) didn't extend top bars to the extent that we do today and this always poses a dilemma.

One response is to redistribute some of the negative moment to positive moment but that only goes so far.

If the rebar is not long enough, then you essentially have a condition where the beam will crack near the ends of the top bars - near the inflection points.
This results in more of a simple span - which is what I sometimes assume. The older beams usually have heavier positive steel than normally required - almost like the original
engineer designed the beams as simple spans and then dropped in some negative steel to sort of control cracking.



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[hdn32]

Thanks JAE

 

[KootK]

Could you clarify your comment "redistributing your negative moments" for me please? I am familiar with ACI's moment redistribution and is this what you meant?

It's certainly related to ACI's moment redistribution method, just taken to the extreme. Here's what you do in more detail:

1) Draw a moment diagram over the span under investigation assuming that it's a simple span. Now you'e got some negative moment capacity that you're not taking advantage of at the supports.

2) Shift your moment diagram upwards until your negative moments cease to fit within the allowable negative moment envelope, presumably based on bar extension limitations on one or both sides of the supports.

3) Calculate an allowable applied load based on the moment diagram in #2.

This is really just a fancier version of treating the slab as simple spanning which JAE and I both suggested above. In theory you can expect some significant cracking over the supports between the points of inflection. This should be discussed with the owner prior to pursuing this strategy.

the slab was cast integrally with transverse beams so there is no rotational displacement allowed and the detailing of the existing slab/beams were never intended for pinned end conditions. Therefore, I am worried that with flexural cracks at face of support/transverse beams due to negative moments, the concrete section area will reduce and it may fail in shear

Yeah, I lose sleep over that too. Does your top steel at least extend a development length past the face of support? If so, I'd be comfortable with the shear as your slab should be in negative bending at the support and dv can be based on the distance from the underside of slab to the top steel.

I once evaluated a one way roof slab spanning over beams. It had no top steel whatsoever. None. And, surprisingly, no evidence of cracking at all. Of course, it sounds as though your slab will be working a good deal harder.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[KootK]

2) Shift your moment diagram

Correction:

2) Shift the ends of your moment diagram upwards.



I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[hdn32]

Thanks KootK