Masonry Shear Wall - Participating Element 1

[CURVEB]

I have a condition where we have 3 north/south masonry shear walls, 2 on each side of a flexible diaphragm, and one in the middle. The one in the middle is short and has openings, therefore is much more flexible than the 2 on each edge. The problem is that due to flexible diaphragm behavior, the middle wall takes a full 1/2 of the tributary width of the diaphragm, but it is the weakest wall. TMS 402 section 1.17.3 states that masonry elements MUST participate in the LFRS unless they are isolated in their own plane. This is a low-seismic building, but has pretty high wind loads due to tall parapets. We are struggling to get the middle wall to work, even as a full-grouted shear wall with horizontal reinforcement. My questions are:
1) Do people generally try to show that the diaphragm meets the criteria of a rigid diaphragm so that the load can be distributed based on stiffness instead of tributary area, even if the diaphragm would be classified as flexible (bare metal roof deck)?
2) Are you aware of any code provisions that allow the wall to be classified as non-participating, even if it is attached to the diaphragm?
3) What manner of isolation is normally used for this condition? It is supporting some gravity loads, so I would think any manner of attaching the diaphragm to the wall would need to be low-friction to prevent transmitting any lateral forces into the wall.

The attached file shows the rough layout of this structure.

Thank you for your thoughts and suggestions.

 

[msquared48]

Use a plate at the top of the center wall with horizontally slotted holes. This will negate its use as a shear wall and allow all the shear to go to the external walls. Just make sure the holes are long enough.

Mike McCann, PE, SE (WA)

 

[CURVEB]

Mike - I think you have the right idea, but I'm wondering if that will actually completely isolate the wall. From a detailing standpoint, I think you also have to separate the wall from the diaphragm on each end, otherwise the diaphragm will "bump into" the wall, putting force on it. On other structures, we've isolated the entire north "wing" area, but I think you could also just isolate this one wall in-plane by putting an expansion joint on each end. Basically just build the wall 1/2" short on each end (as long as the diaphragm deflects less than this amount, no force goes into the wall). Do you think this is necessary or are there issues with this detail? Has anyone ever built it this way before? I think we'll get push-back from architects since the wall will no longer be a nominal masonry dimension.

 

[KootK]

I tend to be quite sceptical when it comes to slotted bolt hole movement joints. Unless they're done all fancy with Teflon bearing pads etc., I think that they're prone to binding up. Here are some other ideas:

1) One path you could take would be to capacity design the wall. Determine an over strength flexural capacity for a nominally reinforced middle wall, make certain that a shear failure can't occur prior to flexural hinging, and let the wall hinge at the base. for the sake of any brittle Finishes / windows, it might be prudent to ensure that flexural hinging doesn't happen under service level wind.

2) You can use ASCE provisions to classify the diaphragm as rigid depending on the ratio of diaphragm deflection to VLFRS drift.

3)You could design everything ignoring the middle wall until the end. Then, design the middle wall for the displacement that you expect for the diaphragm where the middle wall ties in. Use a conservative value of the diaphragm deflection and don't forget to include the drift in the two end walls.

I like option #2 the best.



The greatest trick that bond stress ever pulled was convincing the world it didn't exist.

 

[CELinOttawa]

I agree with Mike and Kootk both; Use Mike's solution as to slotted connections, and specify an appropriate product to address Kootk's concerns.

One example (which I have not personally used, but have been the reviewing engineer for a bridge authority) is:

http://structuralbearingpads.vossen...series-ptfe-steel-slide-bearings-2/item-1026?

Seemed to work well and be simple to install however, again, I was not directly involved in the specification or install.

 

[KootK]

Those low profile bearing pads are slick! We typically spec the assembly ourselves and end up with something as thick as my steel manual. It's overkill for many situations.

Earlier this year, I did some renovation work in a large OWSJ industrial building. Down the middle, along a line of steel beams, the designer had an expansion joint installed. It was unlike anything that I'd seen before. Little 1/4" U-shaped bits of steel were welded over top of all the joists seats with maybe 1/4" clear on all sides. It formed a "tunnel" of sorts through which the joists could slip along their own longitudinal axes.

I didn't like the detail much because of the slack in the direction perpendicular to the joists. Short of installing a PTFE pad, however, I thought that the detail would do a pretty convincing job of allowing longitudinal slip. Not so. The joist seats and supporting beams were painted in the field. It was pretty easy to see that, after forty years in service, no movement had occurred at all. Maybe it just never saw any serious thermal strain or diaphragm loads, I'm not sure.

The greatest trick that bond stress ever pulled was convincing the world it didn't exist.

 

[CELinOttawa]

Aren't they just? I have been keeping them in mind, but haven't yet found the job which really needs them. I try not to look for a job to use a product; Too much psychological chance of a self-fulfilling prophesy. But I do want to use them sometime!

 

[dcarr82775]

Kootk's Option 2 is my first stop when I have this problem.

 

[CURVEB]

Follow up question on Kootk's Option 2 above:

I read the ASCE 7 as stating that the diaphragm can be idealized as rigid if it is concrete or concrete-filled metal deck with a span/depth ratio of 3 or less (with no horizontal irregularities) - section 12.3.1.2 of the 2010. But most of these buildings are bare metal roof decks, so I don't think this applies unless you add concrete to the structure.

The section following is a way to calculate an idealized flexible condition, but I don't read this as saying that if the inequality is not satisfied you get to qualify as a rigid diaphragm. The result of this being that if you cannot idealize the diaphragm as flexible or rigid, you must analyze it using the relative stiffness of both the vertical and horizontal components (semi-rigid condition - as stated in 12.3.1). Does anyone actually do this? Seems like an intensive effort for a small structure.

 

[KootK]

I rarely do that. If I've got a 3-D model going for other reasons or I really think there's a lot to be gained, then I'll do it. Otherwise, it just eats up too much fee. Additionally, I'm quite skeptical of untopped decks as rigid diaphragms no matter what the equalities say. That's an issue for another thread however.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[CURVEB]

No arguments there KootK. Thanks for the input.