Masonry Lateral Ties In Compression

[BadgerEngineer]

I have some questions regarding masonry ties for vertical reinforcement in compression.

Per ACI 530-05: Section 3.1.8.3. “The compressive resistance of steel reinforcement shall be neglected unless lateral reinforcement is provided in compliance with requirements of Section 2.1.6.5”

Per subsection a, b, c of 2.1.6.5
A.) Ties must be minimum 1/4" in diameter
B.) Vertical spacing of lateral ties shall not exceed 16 longitudinal bar diameters, 48 lateral tie bar or wire diameter, or least cross sectional dimension of member.
C.) Ties must be arranged so every corner bar and alternate longitudinal bar shall have lateral support provided by the corner of a lateral tie with an angle not more than 135 degrees.

Typically we use 9 gage Ladder type reinforcement (.144” diameter) at 16” o.c. spacing vertically. If we have vertical rebar in a wall spaced at 24" or 32", etc, we assume the ladder reinforcement acts as the lateral tie for compression reinforcement. We have never run into a problem with this.

However, I realize 9 gage Ladder reinforcement does not meet minimum 1/4" diameter requirement. Ladder reinforcement also does not allow for vert bar to fall in ‘corner’ of tie. I'm assuming per subsection ‘b’ ties must be spaced at a max of least cross sectional dimension –for CMU this is 8”.

Does this mean if we are using vertical rebar in compression for a CMU wall – ladder reinforcement at 16” o.c is not acceptable to act as a tie.

Also, if vertical rebar is not adequately tied per section 2.1.6.5 must you consider the cmu wall/column unreinforced in compression? If yes, not only does this eliminate the fy*As portion of your Pn equation, but it also means your phi factor for axial loads decreased from .90 to .60. (This is per section 3.1.4.1 which states unreincforced walls should use a phi factor of .6 for axial) This will greatly reduce your axial capacity.

Has anyone else run into this? How are you tying vertical rebar in masonry walls for compression if lateral tie must be 1/4" and must fall in "corner"?

This all seems a little conservative to me. What are your thoughts?

 

[civilperson]

Ladder wire installed in the joints does not allow reinforcement to be used for compression. Compression reinforcement must be confined as you described, (with the ties touching the bars to stop outward flexing of the compressive bars).

 

[BadgerEngineer]

So if you have a #5 vertical rebar at 48" o.c. for compression reinf. what is the proper way to tie these bars? You need to use at least a #2, but if there is only one vertical bar in every 6th cell (48" o.c.) - what are you tying the #5 vert bar to? (Possibly the next vert rebar 48" down the wall?)

It has to fall within a 'corner' so you can't simply put a #2 hoop around the one bar.

Possibly I'm not entirely understanding the concept of lateral ties for compression. Is the purpose to stop blowout similar to ties in concrete columns?

If yes, this bending/blow-out would most likely occur in the out-of-plane (of wall) direction & it seems like hooking the vert bars together (down the length of the wall) would do very little from stopping the vertical bars from blowing out in this direction.

 

[concretemasonry]

Are you talking about a wall or a column?

Just because a wall may have some compression on it does not mean it has the be designed like a column.

If it is a wall with the usual range on combined flexure and load applied tie are totally unnessary.

 

[hokie66]

For walls, I never count on the reinforcing as acting in compression. It is there for bending and robustness, not compression. Masonry columns are a different matter, and the confining ties are required. As to the phi factor, consider yourself lucky. The Australian Standard 3700 uses .75 for reinforced masonry, dropping to .45 for unreinforced.

 

[DaveAtkins]

I have designed a lot of CMU walls (admittedly, most of them single story), and I have never needed to rely on the reinforcing in compression. I don't think it adds much axial capacity, anyway. I would grout more cells if you need more axial capacity.

DaveAtkins

 

[msquared48]

I agree with the three previous posts.

If you have a CMU column though, you will need ties, and can only place them at 8" on center vertically, the depth of the CMU block, if you use the normal block. However, with an open block, you can create an open column capable of a tighter tie spacing.

Mike McCann
MMC Engineering

 

[concretemasonry]

The most effective way to get higher vertical capacity in a maspnry wall is to specify a higher strength CMU. The mortar strength is not nearly as important.

If it is a small project you could run into availability problems because the CMU business is based on very high volumes. Without a doubt, with a decent sized project this is a minimal cost since adding water to a batch is cheap and effective. When you push the strength requirements up, it takes a little more cement to have the surface area to make the units stable for the dimensional and appearance requirements. I have seen 8500 psi (net compressive strength block) used with 2500 psi to make 2-high prisms giving an f'm of 4800 psi. - Unfortunately, the ASTM C90 requirements are just minimums that have really not been increased for 50 years, despite dramatic manufacturing and control improvements.

Grout, especially higher strengths, is not a real effective method to increase the wall strength, but it does help until you look at the stress distribution in the unit if there is any appreciable flexure in the wall. - In that case, you want the highest strength material where the higher compressive stresses are (outer fibers).

Not all buildings are in the 15 to 30 story loadbearing buildings that have been designed and built world-wide through the years using ACI 530 and earlier remnents.

Forgive the emotion, but I get upset with the inadequacy of current engineers understanding the proper use and benefits of masonry in the U.S. Elsewhere, it is different, unless there are local provincial codes.

Dick

 

[BadgerEngineer]

The interior wall we are looking at is very heavily loaded in compression. It is a 7 story building with 10" plank (2" topping) and trib floor widths of 35'. We are already using high strength 8" block.

The masonry is highly stressed (especially at the first floor). Since these walls are interior - compression is controlling their design.

It was stated above that vertical rebar is not in a wall for compression. However, If you look at the axial compressive strength equation for reinf. masonry (Pn) there are two parts: One for masonry (.80f'm(An-As) and also one for steel (fy*As). Therefore, increasing steel spacing/size will increase your design strength.

Many times for interior, cocentrically loaded masonry walls - I believe compression will control.

ConcreteMasonry - I'm asking more about vertical reinforcing then mortar strength. I agree that increasing your f'm will help in compression. But we don't like to specify over 2500 psi at which point you can increase axial capacity by adding steel.

 

[DaveAtkins]

OK--now I understand your original question. I think lateral ties are only required for columns, not walls. As you stated, how can you have ties in a wall? And yet, the Code gives a formula for using steel in compression in a wall.

DaveAtkins

 

[hokie66]

If it is not confined by ties, it cannot resist compression. But that is just my opinion based on the Australian code rather than the US code.

 

[concretemasonry]

BadgerEngineer -

I assume that you may in Wisconsin. If so, is the current code recognizing the ACI 530 by complete adoption and reference? The Republic of Wisconsin (was DLIHR? bfore reorganizing) had a long history of partially adopting national standards and then writing "local" language that created some interesting problems and challenges.

In Wisconsin, you should be able to get very high strength units for the lower floors from one of the 2 good CMU producers. An f'm of 3000 psi for an ungrouted prism should not be a problem and should be available. Higher f'ms could be available depending on aggregates.

In all the countries (usually using ACI 530 or similar)I have been involved in (37 total), I have never never seen a case where ties were required in a wall. - This includes 20 to 30 story loadbearing buildings in seismic and non-seismic areas and walls that were even 6" thick. - Granted, your loads, spans and geometry may require some different solutions. A 35' contributory load are is not unrealistic especially if this is an apartment structure with exterior walls and one corridor wall carrying the load.

Take a look at the load paths and distribution to make sure the over-all structure. You oviously have this situation at the lowest floor and you can always go to lower strength units for the upper floors. - I have seen 4 different strengths used in one building.

An old friend, Jim Amrhein, was a leading engineer in research and codes for masonry. He had a philosophy of "Do not totally rely on what you cannot see since this can cause construction, inspection and verification problems. - You can always complicate things by adding steel and grout, but you never know if you are actually creating a problem".

Dick

 

[hokie66]

BadgerEngineer,

Your conclusion that a grouted 8" concrete masonry wall does not work at the bottom of your 7 storey building does not surprise me. The 35 ft tributary loading and the relatively heavy floor system means you would have ultimate loading in the range of 50 or 60 kips/lin.ft., which I think is asking too much of block walls, regardless of the code or strength of block.

You may have to use concrete walls, either CIP or precast, at the lower level or highly stressed areas.

 

[BadgerEngineer]

ConcreteMasonry - A few points:

You state that ties are never required in a wall. Are you implying that engineers should use the assumption that rebar gives no added vertical capacity to a wall OR implying engineers should use vertical rebar and disregard ACI 530 tying requirements? I'm assuming ACI has added a fy*As portion of the Pn equation for a reason.

I completely understand your rationale of wanting to increase f'm over adding rebar. However, the client has informed us that 2500 psi block is what they can get from a local supplier that they want to go with. (I, nor this project, are located in WI so I'm not familiar with code adoption in that area)

I'm not trying to argue your point - I just want to understand why ACI 530 requirements are listed the way they are. What type of failure would you expect from a wall with vertical rebar lacking ties? A concrete column will "blowout" the rebar/concrete if the core steel is not properly confined with ties. However, I'm not sure if I would expect to see the same failure for a heavily loaded masonry wall lacking ties.

If you have a masonry column integral with a masonry wall (With a steel beam framing into it for example) would you add ties? Is this any different than a heavily loaded wall?

Hokie - Thanks for the input. The problem we are running into is that the architect is saying he has done this many times in the past with another engineer. Since this is a prototypical hotel - the designs should be very similar. (Plank spanning exterior wall to interior corridor with one common interior bearing wall) I realize that the "Just because another engineer did it - it can be done" is not a good defense - so we want to make sure we evaluate all our options.

Thanks for all the input guys! Like I've said - I'm not trying to argue anyone's point. I just want to understand WHY ACI 530 code language is listed the way it is.

 

[concretemasonry]

Badger -

I assumed you were considering just walls and not a combination of wall and columns, so some comments were not accurate.

Those block that the owner has selected do really limit your your use of engineering knowledge. By 2500 psi, I assume this is net compressive strength determined by ASTM C140. The ASTM C90 specification requires 1900 psi, but there are few CMU producers that can afford to make a block that weak. Because of this, the block that are called "high strength" and are really just older or well cured block if the production methods are reasonably modern (past 20-30 years).

Have any actual prism tests been provided? These may prove the block are actually much stronger than assumed. If you are just looking at an old supplier supplied test report and the conservative table in ACI 530 that use unit strength and mortar type may be much lower than actual.

Sorry about the "Badger/Wisconsin" connection, but I am close to that problem.

Dick

 

[hokie66]

Badger,

One mode of failure of block walls is splitting of the webs, so if bars in the center are not confined, they could be in the split.

I think the provision for compressive reinforcement adding to the axial capacity would really only apply for thicker walls, say 16" thick or possibly 12", where you could use two layers of bars and tie them together. In that case, it acts like a column.