I want to design a built up beam...say a W24x104 with a W16x26 turned horizontally and welded to the top flange (compression flange). Since the W24x104 has both compact flange and web, should I use the equations in Section F4 of the Steel Manual and use the new section properties of the combined section?
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Help with 2005 AISC Manual Chapter F
- Thread starter LannyBudd
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Remember that your section classification will now be based on the b/t ratio of the W16x26 flange plates in pure compression. Compactness of the W24x104 compression flange should be irrelevant as it will be braced from buckling by the web of the W16x26.
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
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- #4
KootK, what you mentioned is exactly what I was trying to get at. What I would really like to figure out is what the Lp for the combined section would be. If I weld the W16x26 to the top flange of the W24x104 with a weld, for talking purposes 3@12 on both sides of the top flange can I assume that the combined section is now fully braced?
Fun. I'm pretty sure that I went down exactly this path myself a few years ago. May I ask what your application/load source is?
While the torsional resistance, and therefore LTB resistance, of the combined section is meaningfully improved by the addition of the two W16x26 flanges, it's still an open section from a torsion perspective and I don't think that you can necessarily assume that the section is fully braced. I think that the most technically correct way to evaluate the member is still as a singly symmetric section as JAE has suggested.
When I did this, it was late in the evening and I was short on noodling time. I took this shortcut which I still feel is valid:
1) I designed the main beam (W24 for you) such that it worked for strength for all failure modes other than LTB. I assumed full compression flange bracing for LTB.
2) I used the combined section properties for deflection. I had a section builder app so this was easy for me.
3) I figured out the axial force in the compression flange of the main beam assuming that the secondary member (W16 for you) was absent. I then checked the secondary member as a column for that same load. I treated that fictional column as though the weak axis and torsional buckling lengths were zero and the strong axis buckling length was the distance between lateral supports for the combined section (the entire beam length in my case).
The sketchy part of this poor man's LTB check is whether or not the fictional column can be considered to be continuously braced for the torsinal buckling mode. While it takes very little to provide that bracing, torsional buckling of the fictional column and LTB of the combined section are similar phenomena.
I find the weld requirements to be of interest in this situation. In particular:
1) Obviously, you have to satisfy VQ/I demands. Additionally, there's a weld demand created by the need for the members to act compositely when they try to LTB buckle together. I don't know how to put a number to that. I've seen methods proposed to deal with this for built up columns but I'm unsure how to translate that to this scenario. I'm confident that the demand would be pretty small compared to VQ/I requirement though.
2) After welding -- and even without welding -- the combined section has a residual stress pattern that doesn't really match anything that the code flexural capacity checks were developed for. I think that this is generally ignored.
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
While the torsional resistance, and therefore LTB resistance, of the combined section is meaningfully improved by the addition of the two W16x26 flanges, it's still an open section from a torsion perspective and I don't think that you can necessarily assume that the section is fully braced. I think that the most technically correct way to evaluate the member is still as a singly symmetric section as JAE has suggested.
When I did this, it was late in the evening and I was short on noodling time. I took this shortcut which I still feel is valid:
1) I designed the main beam (W24 for you) such that it worked for strength for all failure modes other than LTB. I assumed full compression flange bracing for LTB.
2) I used the combined section properties for deflection. I had a section builder app so this was easy for me.
3) I figured out the axial force in the compression flange of the main beam assuming that the secondary member (W16 for you) was absent. I then checked the secondary member as a column for that same load. I treated that fictional column as though the weak axis and torsional buckling lengths were zero and the strong axis buckling length was the distance between lateral supports for the combined section (the entire beam length in my case).
The sketchy part of this poor man's LTB check is whether or not the fictional column can be considered to be continuously braced for the torsinal buckling mode. While it takes very little to provide that bracing, torsional buckling of the fictional column and LTB of the combined section are similar phenomena.
I find the weld requirements to be of interest in this situation. In particular:
1) Obviously, you have to satisfy VQ/I demands. Additionally, there's a weld demand created by the need for the members to act compositely when they try to LTB buckle together. I don't know how to put a number to that. I've seen methods proposed to deal with this for built up columns but I'm unsure how to translate that to this scenario. I'm confident that the demand would be pretty small compared to VQ/I requirement though.
2) After welding -- and even without welding -- the combined section has a residual stress pattern that doesn't really match anything that the code flexural capacity checks were developed for. I think that this is generally ignored.
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
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- #6
The application…Contractor wants to demo steel, blah blah blah, now there is no steel for temporarily support new massive equipment, blah blah blah, existing steel has an unbraced length of 38’-0” and was designed for an unbraced length of 14’-0”.
The application of this beam that I want to design is for pure flexure only, no torsion, applied axial force.
The moment using the ASD load combinations is 850 kip-ft. Per Table 3-10, a W27x161 is good for 702 kip-ft @ Lb=38’-0”. Unfortunately a W27x161 is not a shape that is readily available via warehouse at the project location so my project manager said try a lighter sections, slap a member on the top flange like a carrier girt and it call it fully braced. The reason for using a W24x104 is because this member fully brace get me a moment of about 720 kip-ft however I am not fully convinced by adding another wide flange turn horizontally on the compressing flange can make the top flange fully braced. Lp is a function of the radius of gyration of the compression flange plus one-third of the web area in compression.
What I am unsure about is how to check the LTB for this combined section for the relatively large moment it will see. All the other checks are pretty straight forward (i.e. yielding, FLB TFY, sidesway…blah blah blah).
In a nutshell, the contractor literally demo’d everything plus the kitchen sink, now they don’t have steel to hoist up heavy equipment. Now I need to design a beam with an Lb of 38’-0” for the contractor to use as a hoist beam.
The application of this beam that I want to design is for pure flexure only, no torsion, applied axial force.
The moment using the ASD load combinations is 850 kip-ft. Per Table 3-10, a W27x161 is good for 702 kip-ft @ Lb=38’-0”. Unfortunately a W27x161 is not a shape that is readily available via warehouse at the project location so my project manager said try a lighter sections, slap a member on the top flange like a carrier girt and it call it fully braced. The reason for using a W24x104 is because this member fully brace get me a moment of about 720 kip-ft however I am not fully convinced by adding another wide flange turn horizontally on the compressing flange can make the top flange fully braced. Lp is a function of the radius of gyration of the compression flange plus one-third of the web area in compression.
What I am unsure about is how to check the LTB for this combined section for the relatively large moment it will see. All the other checks are pretty straight forward (i.e. yielding, FLB TFY, sidesway…blah blah blah).
In a nutshell, the contractor literally demo’d everything plus the kitchen sink, now they don’t have steel to hoist up heavy equipment. Now I need to design a beam with an Lb of 38’-0” for the contractor to use as a hoist beam.
It does not take long to run into the limitations of the "cookbook" layout of the current codes. I would not consider it fully braced. As the OP mentioned, I would take the radius of gyration of the compression area of the combined section and from that obtain the allowable bending stress based on that L/Rt.
If this becomes a long term situation, fatigue may require consideration.
Another approach would be to choose sections that result in a combined Iy greater than Ix. That would preclude LTB. Maybe a little truss on top.
Is there no way to get a rotational brace on this thing at mid span? Maybe a channel welded to the top flange and extended to the roof?
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
Another approach would be to choose sections that result in a combined Iy greater than Ix. That would preclude LTB. Maybe a little truss on top.
Is there no way to get a rotational brace on this thing at mid span? Maybe a channel welded to the top flange and extended to the roof?
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
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- #9
This is for an open structure at a large industrial facility. No roof to brace back to because this is the highest elevation. In this sector of Structural Engineering, steel is cheap compared to everything else so spending $20k for two beams that will be used for a couple of hours is not an issue.
If you look at section F4.2 - at the end of section 2 - there is a reference to "I-shapes with channel caps" for LTB.
This is similar to what you are doing with a W16 instead of a channel cap so I would see this as somewhat applicable.
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This is similar to what you are doing with a W16 instead of a channel cap so I would see this as somewhat applicable.
Check out Eng-Tips Forum's Policies here:
faq731-376
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- #11
I did see that provision about the channel cap, which is why I posted the thread. I tend to be very literal in my interpretation of the code. I don't think even a heavy MC will get me to where I need to be plus heavy channels are hard to find at a warehouse.
I would investigate the stiffness and strength of the W16x26 to act as a lateral brace per Appendix 6 of the 2005 (or 2010) AISC Specification. If the W16 meets the requirements, then I think you could consider the W24x104 as being fully braced. You will need to make a decision as to whether the W16 provides nodal or relative bracing. I would say nodal.
Is the hoist load applied to the bottom flange of the W24? If so, then you should gain at least an additional 15% of flexural capacity based on the load being applied below the shear center.
Is the hoist load applied to the bottom flange of the W24? If so, then you should gain at least an additional 15% of flexural capacity based on the load being applied below the shear center.
@Hokie: I considered looking at the W16x26 as simply bracing as well. I rejected that, however, because I think that something important is missed if the upper beam is treated separately from the main beam. The W16x26 is going to pick up a significant amount of compression as detailed. That compression will affect the available strength and stiffness that the W16x26 can offer the W24x104 in a way that will be difficult to quantify.
The only ways that I'd be satisfied with the Appendix 6 checks would be:
1) If they were in addition to other checks that looked at the composite section or;
2) The connection between beams was detailed to permit longitudinal slip between them. Say, intermittent angles used as keepers.
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
The only ways that I'd be satisfied with the Appendix 6 checks would be:
1) If they were in addition to other checks that looked at the composite section or;
2) The connection between beams was detailed to permit longitudinal slip between them. Say, intermittent angles used as keepers.
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
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