exterior masonry wall

[SteelPE]

I have a building that is to be designed using IBC 06 and is 22’ high (to the support) 110mph category B. The client would like to use 8” masonry walls for the exterior of the building. The building is steel framed with the masonry used as a “cladding”. When I gave him the preliminary reinforcing requirements of #6 @ 16” o.c. he was not happy (on top of which I suggested switching to 12” masonry). He gave me the standard "we have never done that before" and "we don't want a conservative design".

In my design I assumed the masonry would span as a simply supported beam from the foundation to the roof diaphragm. Is it possible to size the masonry for 2 way action? Is this done commonly in practice? This is the only way I can reduce the loads on my masonry wall unless I have a fundamental flaw in how I am approaching the design.

 

[msquared48]

Two way action would imply the use, either of vertical masonry pilasters or steel members, even rigid frames, to take the load. Do you have these available in the framing design?

Also, I assume that wind controls over seismic here for you for the wall design?

Mike McCann
MMC Engineering

 

[SteelPE]

I don't believe that seismic is controlling the design. Everything I have done to this point has been preliminary.

We do have vertical columns that support the steel. Right now these columns are spaced at 36’-4” o.c. but I can close the spacing up in order to get 2 way action to work.

 

[jike]

Make sure that you are using appropriate C&C loads with effective wind area of 22^2/3 = 161 sf.

Is it possible to create fixity at the base of the wall with closer spaced dowels and then reduce the main steel at mid-span of the wall? Or is it possible to raise the top of the foundation wall to reduce the height of the masonry wall?

 

[SteelPE]

Jike,

For preliminary calculations I was using 20.4 psf (C&C 100 ft^2). I didn’t figure the exact wind load. At 22’-0” high I am getting a moment of 14810 in-lbs. which, according to my charts, requires #6 at 16” o.c. (or #7 @ 24" o.c.).

Increasing the foundation wall height is not an option at the moment. Fixing the base might be an option but due to frost concerns top of foundation wall to footing is approx 4’-8”… I’m not sure how much fixity we could get out of the system.

I have also considered moving the attachment of the masonry to the underside of steel but this creates more problems than it solves.

 

[jike]

With the req'd lap lengths, I try to stay with #5's in the wall or at most #6 foundation dowels.

 

[JAE]

You could also provide intermediate steel wind girts (horizontal) spanning between columns. This has the effect of increasing your steel sizes, weights, etc. but provides you with two spans of masonry instead of one long span.

 

[JAE]

Using the NCMA Masonry design program - using 20.4 psf wind load and no other applied loads (self weight included) I get #5 @ 24" o.c. to work. Moment shown is 14,757 in-lbs/ft.

 

[SteelPE]

JAE,

That's funny, using a moment value of 14757 in-lbs, my NCMA program tells me I need to use #7 at 24" o.c. or #6 at 16" o.c. or #5 at 8" o.c.

 

[mudflaps]

I have real heart burn considering an 8" CMU wall as cladding. If you went to 12" CMU in a bearing wall condition it wouldn't be cladding and the wind sure wouldn't know the difference. Call the building department and get their thoughts on this. Lots of engineer don't like the current wind design methods. There's gotta be a reason for that.

I like the idea of adding a mid-height wind girt to the structure behind. You could also move the vertical rebar and increase the effective depth. Depending on how much suction you have on the down wind side you might have to add a second layer of smaller verticals.


Old CA SE

 

[spats]

JAE,

I also used the NCMA program, and I didn't get your answer. I understand there is a Version 4.1.3 update (I have 4.1.2), but I can't believe that has anything to do with it. What are you using for "design basis" in the program. For prism strength of 1500 psi (only strength allowed by program), masonry compressive stress governs. I agree with SteelPE's answer.

As an alternate I looked at a masonry flexural table that I also use: it says #5@24 resisting moment is 11,831 in-lbs/ft, well short of the actual moment. #6@16 resisting moment is 15,313 in-lbs/ft.

 

[SteelPE]

Mudflaps

The use of the CMU as cladding was selected in order to facilitate the process and to provide flexibility in the future. Currently 2 additions are planned in the future. I didn’t mention the additions before because I believed it had nothing to do with the question I was asking. These future additions will require the removal of the existing masonry walls which would be difficult if the exterior walls were masonry bearing.

Spats

We must have the same design tables because that is the exact number I have for #6 @ 16” o.c and #5 at 24” o.c.

 

[spats]

mudflaps,

I also have to disagree with your assessment that masonry, load-bearing or not, should not be considered components & cladding for out-of-plane loading. It definitely not MWFRS!

 

[JAE]

I was using 2002 MSJC Allow. Stress design and got #5 @ 24" o.c. to work with the following:

#5 bars
fy = 60 ksi
Depth to bar (x) = 3.8125"
Height of wall = 264"
Wind pressure 20.4 psf.
Partial grout running bond
CMU - 125 pcf density
Portland cement Lime
Type S mortar
Coarse grout
f'm = 1500 psi
no other loads entered.

NCMA Program version 3.1.1.2


When I switch to 2000 IBC Allow Stress method it requires #6 @ 16" o.c. - quite a difference.

 

[akastud]

I sometimes like to use a #5 bar in each face to increase spacing. It usually requires the same amount of steel (sometimes a little less since you divide by d^2) but you use half as much grout.

akastud

 

[JoshComfort]

JAE - is your program allowing a 1/3 stress increase?

 

[JoshComfort]

Sorry about the multiple posts...

The tables I use are in NCMA TEK 14-19A and a #5 @ 24" would only work with the 1/3 increase, which isn't permitted when using the Basic Load Combinations in the 2006 IBC.

One last thought is that you may be able to get #5 @ 24" to work using Strength Design.

 

[SteelPE]

According to NCMA 4.1 strength design is the answer.

Here are some quick numbers:

For ASD #5 @ 24 provide Ma = 11,831 in-kips.

For Strength Design #5 @ 24 provide phi-Mn = 28,329 in-kips.

Divide this by 1.6 to go from strength design to allowable design in the load combinations and you get 17,706 in-kips. A 50% increase in the strength. I wonder if this is because ACI 530 allows a 1/3 stress increase and IBC doesn’t.

There definitely seems to be a problem here. I wouldn't suspect a large difference between the two methods.

 

[JAE]

JoshComfort,

Yes I think that is the difference between MSJC and IBC.

Here's the commentary/help in the NCMA program version that I have:

[blue]Choose from the list of six possibilities of design code. Two choices of building code are available, MSJC and IBC. The MSJC Code includes the 1995, 1999 and 2002 issues. The IBC 2000 code includes both allowable stress design and strength design methodologies. The allowable stress design method is based on the 1999 MSJC Code with modifications in the required loading combinations. The 1/3 increase in allowable stress for loading combinations that include wind or earthquake are not permitted in favor of using a load factor of 0.75 when two or more transients occur simultaneously. IBC Strength design uses its own load factors except when the loads include F, H, P or T. In this case load factors from ASCE 7-98 are to be used. The option to use a 1/3 stress increase is available only when using allowable stress design and when the “compute using load data” option is not selected. If section forces are calculated using the “compute using load data” option, the software applies the 1/3 increase to appropriate load combinations.[/blue]

So that explains it - it was applying a 1/3 stress increase automatically and when I switched to the IBC it negated it as the IBC doesn't allow it.