ASD & Post-Installed Anchors 2

[BSVBD]

(If interested, i have a semi-related post entitled, "ASD to LRFD" at "Where is Engineering Going In The Next Five Years". Although the following post is related to "Where (structural) Engineering is Going...", since the following post is technically, mostly structural, i figured i should post it here.)

I am ASD only and would like to stay that way. However, if I cannot, then I will move on and make the necessary transition.

HOWEVER... Is there any feasible way to maintain ASD when sizing and specifying post-installed anchors?

It seems as though Powers and Simpson are trending toward offering ONLY comprehensive LRFD tables for some of their products in their catalogs. Therefore, when attempting to refer to ASD tables and discovering that some prerequisite design criteria data, such as spacing and edge distances are not provided, the tech support for the proprietor tells me that "the company" is desiring to persuade us toward the LRFD trend, which requires us to engage in the electronic program.

The ASD method isn't broke! Why are we trying to fix it? Worse yet, why are we trying to omit it?

Other than using older published data, if the product remains available and the nomenclature doesn't change, Is there any feasible way to maintain ASD when sizing and specifying post-installed anchors?

If anyone thinks that the LRFD pursuit is REALLY for the greater benefit of all, i'm willing to listen and consider. You may, then, wish to provide input within the "5 Years" post.

Thank you!

 

[XR250]

Ditto on ASD. Serviceability controls most of the stuff I do anyway.
I only got about another 10 years in this business so hopefully I can hang on.
I just multiply the loads by 1.6 and use the LRFD tables when needed.

 

[hawkaz]

I prefer ASD also- but I think this boat sailed years ago (at least for ACI 318 Appendix D applications)
I think we are stuck with LRFD

When I was in school in the 80s, they only taught Steel design using LRFD- because it was the wave of the future (Still have my AISC LRFD- 1st edition). I learned ASD on the job- and never go back voluntarily.

 

[TehMightyEngineer]

I posted in your other topic so I'll not repeat that here. Below is mostly my opinion; take it as such. I'm sure I don't speak for everyone.

For ASD on anchor bolts I've used either. The big problem is, as anyone who has had to really dive into Appendix D on ACI 318 knows, anchor bolts and strength of dowels in concrete is not simple and is highly sensitive to installer proficiency and attention to detail. Consequently, to ensure safe capacities the anchor strength is highly penalized with a big safety factor (or phi factor, whatever).

Then, when we run our numbers and get a huge anchor bolts and the contractor rightly states that a "1/2 bolt can easily hold this!" we need to have all sorts of rules on spacing, edge distance, transverse reinforcement, etc. to get the absolute maximum out of our anchor bolts that we safely can.

Your problem to me is two-fold. First, the anchor bolts per ASD are fine but switching to LRFD is simple enough; just moving the safety factor from capacity to the demand side of the equation. However, it is an additional step in design so I can see the annoyance. The other problem is Appendix D of ACI is likely governing your "LRFD" anchor bolts, and ACI is LRFD only (more or less). Anchor designers know this and are more and more working based on Appendix D from what I've seen. They'll give ASD capacities in their catalogs so you can get a quick number from a simple table in a pinch, but for the real "engineered" capacity you need to consider Appendix D, ACI 318, LRFD, and so on. As the Appendix D numbers are tedious to calculate computers become very handy and often engineers will spec fasteners based on whichever computer program gave them an easier time calculating the Appendix D capacities. Thus, you can run the LRFD numbers from the published tables and factors but the manufacturers are going to recommend their computer program for the really complicated details (and above all, a computer can't do everything. I do a lot of anchor bolt work and regularly need to run hand calculations for when I'm outside what their canned programs can do; or worse I find an error in their calcs.)

All said, don't hate LRFD; hate appendix D. Actually, don't hate Appendix D (except for the seismic parts); hate installers who can't be trusted to clean a hole or give you the proper embedment. It's not ASD vs LRFD; it's that our old way of calculating anchor bolt capacities wasn't good enough. It just so happens that our "better" ways for calculating these capacities are in the codes that use LRFD only.

Professional and Structural Engineer (ME, NH, MA)
American Concrete Industries

http://www.americanconcrete.com

 

[Ingenuity]

Imagine if you had to do LRFD AND the metric system?

How does the rest of the world cope?

 

[TehMightyEngineer]

Ingenuity: "Back in my day our anchors could hold MCCCXXXVII sthène and worked perfectly; you young kids with your fancy math!"

</sarcasm> No offense to the old timers meant; just because a method is old doesn't mean it's bad. That said, beware getting stuck being that person who states emphatically that "I've done it that way for 30 years with no issues!" Maybe so, maybe not. Yes, we used to design buildings just fine using ASD and "old" methods. That said, newer buildings are built to be lighter, cheaper, etc.; the older methods may not get us what modern owners and contractors expect. Plus, we've learned what areas we had excess capacity in and what areas we did not have enough capacity in (special steel moment frames in high seismic regions being the best case study of "done it this way for X years, darnit!" in my opinion).

As a (hopefully) balanced, younger engineer I try to recognize when a simple "old-fashioned" approach is best and when I should use the most modern approach.

What I think you're really having the biggest issue with (and I totally agree with this if you are) is engineers are being forced to spend more and more time doing refined, advanced analyses to keep costs down for more complex structures. Yet, again to keep costs down, we have to have competitive fees. Thus, we're forced to save other people more money at the expense of our time while being asked to charge the same rate we charged 20 years ago. This is the issue; and one we as engineers need to address (in another topic).

Professional and Structural Engineer (ME, NH, MA)
American Concrete Industries

http://www.americanconcrete.com

 

[wannabeSE]

The ICC ESR for post installed anchors typically have a procedure for ASD using a weighted average of the governing load combinations to convert the capacity. But, I find it easier just to use LRFD. If you are gaming the system with seismic loads, this might be used to take advantage of the 20% capacity increase for ASD load combinations with the overstrength factor per ASCE 7-10 §12.4.3.3.

 

[rn14]

On a side note, something that kills me is that for seismic, if you cant guarantee ductile yielding in the steel you have to use omega. That seems completely arbitrary to me. I cant seem to see a correlation between a building systems omega factor and anchor bolts.

 

[sandman21]

Our office plans on going back to using slide rules and WSD.

 

[MacGruber22]

...engineers are being forced to spend more and more time doing refined, advanced analyses to keep costs down for more complex structures. Yet, again to keep costs down, we have to have competitive fees.

I definitely see this problem occurring. That said, I don't really subscribe to the idea that LRFD contributes this issue. Both ASD and LRFD have load combinations with some sort of capacity check. With respect to concrete anchoring, Hilti and Powers offer very powerful software for free, and rarely do I ever have a situation that is outside their ability. I will say that, initially, it may be a bit disconcerting that it is difficult to fully understand and remember ACI's App D equations, but the ACI chapter isn't voodoo magic. Once you have a feel for the limit states, the computation is much easier to perform with confidence.

Plus, there are many ways to provide reinforcing to avoid the concrete capacity limit state checks, which leaves you with only checking the anchor steel and minimum embedment to preclude pull out. Easy peasy.

"It is imperative Cunth doesn't get his hands on those codes."

 

[BUGGAR]

I have learned how to make most of my concrete anchors thru-bolts. Now I can spend more time on more critical life safety issues.

 

[MrHershey]

On a side note, something that kills me is that for seismic, if you cant guarantee ductile yielding in the steel you have to use omega. That seems completely arbitrary to me. I cant seem to see a correlation between a building systems omega factor and anchor bolts.

We could certainly debate whether the building structural system's omega factor is correct or if we should be using some sort of component-specific omega factor. The concrete code isn't 100% clear but I'd argue (and people much smarter than me would appear to agree) that we should be using a component-level omega factor for anything governed by Chapter 13. Component level overstrength factors weren't included in ASCE 7-05 or the first two printings of ASCE 7-10. Were not released until 2013 in Supplement 1 (PDF link). Would note that most of the component-level overstrengths are the same as your normal shear wall building overstrength. System overstrength really doesn't vary that much in the grand scheme of things. All of Table 12.2-1 is between 2 and 3, for all but cantilever column systems. Your overstrength in ordinary plain masonry walls (R=1.5) is the exact same as your overstrength in buckling restrained braced frames with moment resisting beam column connections (R=8). NEHRP has some great discussion on what goes into the overstrength factor in FEMA P-750.

But in any event, this requirement is a good one and one that is frequently missed. Can argue about the merits of using this omega factor versus that omega factor, but the general concept is far from arbitrary. Breakout and pullout failures are largely governed by the concrete material strength. It'll just pop or fail abruptly on you, there won't be much yielding or warning before it goes. It's the same idea as making sure moment frame beams fail in flexure instead of shear.