Anchoring baseplates into slab on grade? 4

[mfstructural]

Hey all,
I have a situation where we are adding a mezzanine to an existing warehouse building. The warehouse is about 30' high and the mezzanine will be about 10' off the slab. I am working with another structural engineer and architect on this project.

The mezzanine is planned to be 100' x 40', with 3 columns per frame. There will be 8 frames spaced at about 12'.

My issue is that they want to anchor the column base plates directly to the slab on grade which is 6" thick. I had suggested providing footings but they claim anchoring to the slab is sufficient. They are also proposing to have the columns designed as flag poles and this would be the lateral system...no bracing or moment frames. The building is in northern Michigan so not a high seismic area but I'm not sure I like the idea of anchoring directly to an existing slab.

In the past I've always provided footings. I suppose if you show the anchors are sufficient with a sufficient amount of clearance between the bottom of slab and end of anchor hole and also that the slab can resist the moment (assuming no reinforcement) then this would be feasible. The mezzanine will be used as an office so it's 50 psf LL.

We are planning on spanning 2x12's or wood trusses between the frames so the DL will be low. I'm not sure if this will be feasible; the architect is looking into fire protection requirements.

Any thoughts or has anyone ever anchored to an existing slab on grade?

 

[MrHershey]

Depends on the load level. You'll need to check that the slab can handle it. To do this you'll need some subgrade information from a geotechnical engineer. I'm not aware of any assumptions that can be made for subgrade modulus in the absence of geotechnical information, but perhaps others would have a reference.

Given the lack of information, I would tend to argue for a small footing.

 

[JAE]

I would not do it.

1. The slab-on-grade as "6 inches" may not be exactly 6 inches thick. Many times slabs vary in thickness due to the variation in subgrade.
2. You say that the slab is unreinforced. If so, you would need to check punching shear and have a thick and wide enough base plate.
3. With an unreinforced slab, and a cantilevered column, how are you resisting the column base moment? The slab has no reinforcement and this has very little (if any) moment capacity.
4. Mezzanines inside in low seismic areas still need to resist 0.01 times the dead weight of the mezzanine laterally.
5. Also - you technically could argue that any lower and upper walls on the mezzanine could experience some air/pressure differential due to exterior wind forces (i.e. the lateral 5 psf required in the IBC for example). So your lateral could be significant.
6. ACI recommends a minimum of 6" thickness for footings. If your slab is just at or below 6" you would be in violation of that provision.



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[jike]

You should be able to prove whether the slab on grade is adequate or not. If the DL = 20 psf and the LL = 50 psf then 12.5 x 13.33 x 70 psf = 11.7 kips. If you have partitions, you might use 80 or 90 psf total which could be as high as 15.0 kips. Seems like too much for a 6" slab. I suspect that the slab bending and punching shear stresses would exceed allowables but you need to do the numbers. With lateral loads, stresses will be worse.

I suggest using individual footings.

 

[mfstructural]

I am thinking the same thing. For some reason the engineer that I'm working on this with wants to bear on the slab. I'm thinking we can cut out a 2'x2' opening in slab and provide a thickened slab of 12" and dowel into the existing slab. I'll run the numbers within the next couple days and come back with the results.

I'm not licensed in the state the building is being erected in so I'm just on board to prepare the calc. I'm only a PE, SE in Illinois. regardless of sealing it or not, I can't agree to do something like this if the numbers or my judgment is against it.

 

[msquared48]

JAE brings up some good points. Listen to him.

I have done this before, but only for a wood stud bearing wall. Any isolated column or column at the end of a beam I have opted to saw cut the slab and install a new spread footing - no exceptions unless you are dealing with a minimum 8" slab with two matts of reinforcing, not an unreinforced 6" slab.

You mentioned cantilevering the column off the slab for the lateral. You may be able to do this in theory for steel, but not for wood columns. Personally, I would use wood shear walls or steel moment frames, but the Architect must be a better structural engineer than me.

Mike McCann
MMC Engineering

 

[mfstructural]

All good points. They don't want shear walls or bracing because they want to keep the area "open". That's why he came up with this. Using moment frames is a good option because you will reduce the moment at the base of the column/footing. It's different if this were a storage rack but this is a structure and people will be on it often. I don't like the idea of flag poles as lateral system.

 

[WillisV]

Also note that cantilevered column systems do not qualify as an "R=3" system (some people have the misconception that once you are in SDC B or C you can do whatever you want as an R=3, which is true to some extent, but not for cantilver/flagpole systems) so even in a seismic design Category "B" you would need to use a "Steel Ordinary Cantilever Column System (OCCS)" with an R of 1.25 and designed per the OCCS provisions of AISC 341 which basically amount to you must keep axial demand below 15% of axial strength.

 

[ztengguy]

Just curious, why cant the engineer signing and sealing the job do the calcs and leave you out of it?

Run your calcs, provide to the EOR, and if he chooses to ignore calcs...talk to the owner.

 

[Jerehmy]

I would switch to steel/aluminum and use moment frames pinned at the bottom.

Hilti has adhesive anchors that are quite strong and don't have to go very deep. 1/2" diameter anchor embedded 3.5 inches into normal weight concrete has an allowable tension capacity of 3kip and allowable shear of 3.7k based on bond strength as an example.

I'd also try and keep the leveling grout to a minimum.

Is the mezzanine going to be near any exterior walls/columns? I'm guessing this is a prefab metal building? You could just tie to them for lateral support.

What about slab reinforcing. I have to believe it has something. I'd determine a safe allowable load for the slab and space the columns accordingly. Tie the mezzanine floor to some exterior walls for lateral bracing so you won't have to worry about high overturning base plate loads and/or switch to steel/aluminum and use moment frames.

Why does fire protection make it not feasible and why is there fire protection even needed? It's not a separate structure, it's inside an existing one. The same people that use the warehouse will be using the office space, no?

 

[Jerehmy]

I want to clarify, by fire protection I meant separation from the warehouse with gypsum, not sprinkler systems, etc.

 

[TehMightyEngineer]

What WillisV said. However, note that ASCE 7 section 11.2 lists the definition of a cantilevered column system as "a seismic force resisting system in which lateral forces are resisted entirely by columns acting as cantilevers from the base." Thus, if you use moment frames in combination with the cantilevered column then you do not have a R=1.25 system.

Still, much better to just pin the bases and use moment frames. Putting moment into a minimally reinforced slab-on-grade seems like a terribly inefficient design to me. Also, the large baseplates required to transfer such a moment may present a tripping hazard.

Maine EIT, Civil/Structural.

 

[mfstructural]

ztengguy....I have no idea why he wants me to prepare the calculation. It's a small enough job that it's not necessary but I got the feeling he just wants a peer review/prep process.

Initially he didn't want to tie off to the existing building, but yes it's a prefab warehouse structure. I'm assuming there is some sort of reinforcing in the slab just not sure yet. I'm supposed to get more info tomorrow.

I don't know much about fire protection. I'm still waiting on information. I don't think it should matter because like you said, you're inside an existing building. It's not a new independent structure.

I agree with Teh; no reason to introduce large moments into the slab. So I'm going to go a different route and go with the moment frames.

Thanks all

 

[Ron]

What JAE said...

 

[dik]

I've done numerous alterations of this ilk... and unless there are overturning issues, I wouldn't normally dowel it into the existing slab... I typically use 12" thick... very little additional concrete and often less than the normal minimum concrete truck load for a bunch of them.

Dik

 

[JLNJ]

You can run the calcs and see if you have the required strength, cantilevered or not. JAE is, of course, right about the slab not having much overturning resistance.

One of the items I struggle with in a case like this is the location of the control joints or cracks in the slab. You need to make sure whatever assumption you use to check your design works with the joints and cracks in the vicinity of your posts. Sometimes this is not known and gives me great pause.

That said, there are many, many heavily loaded racking systems which are set on warehouse slabs throughout the country to no ill effect. The architect is right to push you a bit on this one.

 

[msquared48]

On one occasion with similar restraints, I extended the footing on top of the slab, and under the new baseplate. Looked like _(&%, but it worked.

Mike McCann
MMC Engineering

 

[TehMightyEngineer]

Mike, not sure what you mean by "extended the footing". Are you saying you built a footing on top of a slab?

Maine EIT, Civil/Structural.

 

[msquared48]

Yes.

Mike McCann
MMC Engineering

 

[JAE]

JLNJ - I'd agree with you on the fact that there are probably many mezzanines on slabs. But not necessarily cantilevered column type mezzanines.
I think the vast majority are probably braced system (x-braces) or some kind of semi-moment frame system - or tied into the adjacent structure.



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