Analysis Discussion: HSS Baseplate, usually assumed as pinned. Yet in my FE analysis it seems Fixed. 2

[IngDod]

Greetings, attached is a summary of my FEA analysis.

I am currently designing a two story OMF building. My initial assumption was to assumed the columns pinned at the base, I did this because it seems to be the common practice. But now that the time has come to design the base plates I'm having problems (cant sleep) simply assuming its pinned and designing it solely for axial force and shear (using AISC's DG1), it troubles me that if this connection is actually fixed (or at least with high rigidity) the baseplate will fail.

Now, I've always had problems grasping the concept of steel connection design; some people say that since you assume is pinned and you designed for axial loads only then it will behave as such. I proceeded to do a FE model, if the connection is indeed pinned then I should see little (no connection is truly pinned of course) moment being transferred to the baseplate. What I get is that the base-plate bends considerably due to the moment, now i would expect this since it is precisely this bending which allows rotation; however the stresses in the plate are much higher than what it can resist... So while it rotates, it seems to me that it would fail too. Now i proceed to increase the baseplate size so that it can resist the stresses, now the rotation is much much smaller... giving me the impression that by increasing the thickness I have made the connection rigid.

Please see the attached summary as it is much easier to understand. The initial baseplate designed for axial load only is 1cm thick, while the one designed based on the FEA is 2.3cm thick. I added results for more thicknesses to study the behavior, with 5cm thickness the rotation seems very low.

I don't really know how to judge if the connection rotation is high or low, or whether it approaches rigid or pinned behavior. I'm getting rotations of 2 degrees for the 1cm thick plate and around 0.5 degrees for others... is this a lot? is this rigid?. How much should a connection rotate in order to be considered as pinned?

Thanks and I hope you can help me with this dilemma.

 

[AELLC]

I have done steel OMF in the past, and I always provided a conservatively thick baseplate without worrying at all about the partial fixity issue.

In fact, one was 11 stories tall. and was built about 1986. It is fine to this day. We did not have FEA on computers back then where I worked. We had a simple 2D frame analyzer that ran on a strange "micro-computer".

 

[AELLC]

About the same time I did a one-story OMF, but the columns were about 20-24 feet tall. The drift was not good, and I provided a full fixed base plate. Like I said, we had no FEA and little Code guidance so it made liberal use of stiffeners-

 

[AELLC]

If you run the frame with pinned base condition, you are probably going to see only a very small rotation there at the base of the columns, and that isn't trying to stress that baseplate as much as you may think.

If you run it as 100% fixed base, than your moment maxes out as the rotation is zero.

Now when you run it FEA, that may create a false impression of localized high stresses because I am assuming this a linear FEA analysis. Also, the more you thicken the base plate, then you are increasing the bending moment in the base plate as it approaches that theoretical 100% fixed condition.

 

[IngDod]

Thanks for your reply, I feel that sometimes all the codes requirements and load factors are an overkill; so many building went up before all this fancy design methods and most of them are still standing. But still, it nags at me not to know how it actually works... Cause if the base-plate does take moment it will go straight to the footing... and moment at a footing not sized for moment could be pretty bad.. specially during an earthquake.

I found this classification system for connections and would like to hear your thougts:

Link: Starts at bottom of page 46

http://books.google.co.ve/books?id=...#v=onepage&q=low connection stiffness&f=false

It says that if the connection moment capacity is lower than 25% of the fixed end moment capacity then the connection is pinned. How could I apply this to my baseplate situation? basically i would like to come up with a criteria that allows me to say if the connection is pinned or not.

I supposed i could model the frame with fixed restraint at the column base.. this would give me the fixed end moment capacity when i perform the analysis. Here is the part i dont understand... say I design the baseplate to have 0 moment capacity.. very thin plate... according to this it would be pinned... yet when the structure is loaded would the plate not yield and tear due to the moment.. and should a FE analysis not show you no moment at the baseplate?

Thanks.

 

[IngDod]

Im using non-linear analysis in SAP2000. I have very high spikes in stress near the anchor bolts but otherwise the values seem agreeable.

My problem is that.. true i can run the analysis with pinned ends.. i get no moment at the base.. then i detail the baseplate with four bolts and the thickness for the axial load only... is this supposed to make it so that the baseplate takes no moment in real life?

 

[AELLC]

The moment gets bigger as you make the baseplate thicker, but if the footing compresses the soil a little bit as it starts to rotate, than the moment decreases.

You are overthinking this. If you had only the AISC code book, your frame designer software, and no FEA, you would be OK with this.

Besides, the rotations we are talking about are very small. The columns bases are not rotating very much according to your output, right?

 

[CELinOttawa]

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I feel that sometimes all the codes requirements and load factors are an overkill; so many building went up before all this fancy design methods and most of them are still standing.

Be careful... Modern codes and methodology have allowed our modern world's construction methods. Very few buildings have survived from before modern codes (gravity codes in 1910s & 20, seismic with a reasonable middle step from the 30s to the 60s). I do a great deal of work on historical structures, and not a one could be built today without losing your shirt. There is a good reason for that: premium materials, expertly installed and finished BUT with tons of redundancy because so few knew what was really going on.

Shake your head, apply the codes correctly and fight to correct them when they become unnecessarily complex.

 

[IngDod]

@AELLC: So in your opinion i should treat this as pinned and design the baseplate and foundations for axial and shear load only? Have you designed buildings with HSS and baseplates similar to the one in my file and assumed pinned? (four bolts, one at each corner) if you have it would definitely make me more comfortable in proceeding with my design. I ask because there is little literature regarding hss column baseplates, if it were a W section I would not be conflicted as there is ample researh in to the behavior of this connections.. basically you put the bolts between the flanges and this results in plenty of rotation.

@CELinOttawa: Thanks; I might complain a lot about the codes and curse them for the many complexities they bring but I always follow them to the letter.

While right now im inclined to designed the base plates as pinned, I still would like to hear more from the community in the matter... specially regarding classification using the results i obtained.

 

[Ron]

The primary reason most buildings still stand after many years is that they have never seen the loads for which they were designed. When buildings are subjected to near their design loadings, we see more failures.

As for your model, if you have the capability in your FEA program, animate the model. This will show you how the model is acting under load and you can tell if you have a modeling issue.

Just because you model something in a particular way does not mean that it will act that way in the field. Model it the way you expect it to react....don't model it then expect it to react to your model.

 

[AELLC]

I have designed plenty of buildings with baseplates having the anchor bolts in the 4 corners instead of being "tucked-in" between the flanges of a W-column. No problems.

I am just wondering why you refer to the baseplate as HSS, and the printout says it is A36.

Also, why are you using a square tube for the column? Is it large enough? - I did notice a very large rotation in two cases.

 

[SAIL3]

the first cut from an FEA program gives one a fairly accurate initial distribution of stress based on the initial stiffeness of the various components....what it may not capture is the redistribution of that stress as portions begin to yield...that is why one should have a solid grasp of the underlying engineering principles involved so you can make an independent engineering judgement on the program output. In reality, to design a truly fixed base pl condition would take a dedicated effort as there is usually enough "give" in the fdn as awhole to render the condition pinned. I usually design HSS base plates as pinned, however, if I am using post-instaled AB's , I design the AB's themselves for a fixed condition. My advise would be to turn the FEA progam off and analyze it as pinned and use a pracical base pl thickeness. For a typical stucture, I would not use anything less than 1/2" for base pl thickeness and would tend to go to 3/4" as a min.

 

[CELinOttawa]

As AELLC and SAIL are saying, the stiffness of the model is a moment in time... It does not reflect the changing conditions of reality, and maybe never will.

Frankly every engineer should read "Wooten's Third Law" five times before being allowed to touch a computer.

I never design exclusively by PC, and I would not have looked twice at the output you're stressing over. That's *not* an insult; I am setting you a challenge: Look beyond the output, after verifying there are no errors. Every design I put out is checked to within 20% "back of the envelope" style before I ever believe the computer or sign and seal.

 

[AELLC]

As far as the Code being "overkill" - I have no problem with that.

I recently received news that the floor in a wood residential structure that I had designed long ago was finished with 3/4" mortar setting bed and natural stone flooring - had never heard of that during the design. It weighed a lot more than the ceramic/porcelain/travertine flooring that is commonly used here.

Also we get a lot of instances where the roofing is changed from plain concrete tile to clay tile with buttered end joints (without notification to me) - I actually heard from a builder that a roof over a porte-cochere sagged noticeably, and he went in and installed larger beams without any legal action taken against me, so I was lucky that time, also.

 

[Gumpmaster]

I second the requirement to read Wooten's law.

If you design it as pinned, but some moment is actually transferred to the base plate, what happens? If you base plate yields is that a bad thing? I would say no. Yielding a flat plate in the weak axis is very predictable and ductile. You can do it many times for small strains without adverse effect. Sail3's recommendations are good. Base plate yielding is fine, anchor bolt failure is bad.

 

[CELinOttawa]

Timber is a pretty large exception to most of the common structural rules, particularly under the US and Canadian codes. The majority of structures are so far above the code minimum due to the 5% rule applied in timber design. This is necessitated by the fact that timber is such a variable material, but stress grading is brining the inherent redudancy of timber down a great deal.

I think we will start seeing timber collapses of the nature of the occassional steel and reinforced concrete collapse we see now, but not for another thirty or so years (20 at an absolute minimum).

 

[AELLC]

I think SAIL and Gumpmaster concisely nailed it.

 

[AELLC]

CEL-

I never head of Wooten...that was interesting.

I have read Russell Fling's (Designer of 30,000 concrete beams by ASD elastic method) essay of why USD design is nonsense. That was good read. (But I lost my copy)

 

[rowingengineer]

There have been a few post on this forum about semi rigid analysis of base plates for portal or similar frames, I prefer this method.

This allows you to estimate the defections a little better and also predict the moment distribution due to your detailing a touch better.

http://www.nceng.com.au/

"Programming today is a race between software engineers striving to build bigger and better idiot-proof programs, and the Universe trying to produce bigger and better idiots. So far, the Universe is winning."

 

[CELinOttawa]

USD? Do you mean Limit States Design (LSD), the trip which most every Structural Engineer "takes" daily? I'd be very keen to read that, and either my google fu is down or it is hard to find on the net...

Oh, and frankly Wooten is the man. I mention the third law weekly, and daily if not hourly when training an intern.