Further development from recent court proceedings: https://www.nbcmiami.com/investigations/Engineers-...

I find the statement about conflicting meeting minutes troubling:

But FIGG has challenged the accuracy of the minutes of the meeting, prepared by BP&A, and six weeks after the collapse provided its own version of the meeting to the National Transportation Safety Board, calling them "corrected minutes."

FIGG's version made "significant substantive changes" from the original minutes, according to a lawyer for The Louis Berger Group, an engineering firm also being sued in the civil cases.

I would think that all professional engineering companies require the chair of the meeting to send out a draft of the meeting minutes to all attendees for review and update with any comments or missing information. The final draft isn't issued until everyone agrees, so I find it hard to believe "corrected" meeting minutes issued 6 weeks after the collapse to be considered valid once checked against e-mail records provided by the parties involved.

This also highlights the importance of documentation in our profession.

I don't know how they did it there, but most of the meetings I've ever been in, any minutes were pretty rudimentary and were not emailed around until everyone agreed, etc. So I wouldn't doubt that the minutes were incomplete or omitted items that people said, etc., unless they purported to be otherwise.

I don't think the Minutes of the Meeting on the day prior to the collapse carried much weight in the case.

FIGG is the designer. Only the designer has the right to modify or repair the bridge. The designer is also charged to evaluate the severity of the cracks and their impact to the structure. He is the sole party to decide if a repair is needed, when and how it is to be executed.

FIGG proposed to re-tension the truss members. The MCM contractor was executing FIGG's repair scheme and FIGG knew traffic was running underneath and did nothing to halt the remedial work. The bridge collapsed "during" the execution of FIGG's repair scheme.

Whatever written on the Minutes can't change the above.

The NTSB published photos, that shows a tape can be inserted to a depth of 4" and then 6" into the cracks of the slab and Member 11 respectively, were taken by the workmen on 13th and 14th so the participants to the meeting on 15th would know full well the severity of the cracking. If FIGG concluded, there is "no safety concern" as recorded in the voice mail then that is enough to demonstrate FIGG's competency in this project.

Unless the minutes of the meeting were a condensed version of a recording, it is "he said, she said". There were probably notes taken during the meeting, but the minutes would have been prepared and distributed later, presumably after the collapse, so you would have to question their reliability, just as you would question "corrections" made six weeks later.

in order to find core reasons and systemic failures in procedures the true course of actions / nonactions needs to be established
this is beyond the who to blame level,this would save life / health in future

casting of a shadow on things means resources / thoughts are bent elsewhere
engineering maturity is not making the beeline off of the site, not even mentally. but full support to uncover all that may be found out.

in this case, the fact is evident that certain meeting discussions / decisions need to be cast into mutual agreement BEFORE further work is executed.
professionally this should be mandatory for security related issues.

Roland Heilmann

I thought I had read that the work crew were already up on the bridge while the meeting was taking place. It raises the question whether they on site to perform work based on the smaller cracks and somehow the word wasn't circulated that a re-think was in the works.

Obviously, the meeting minutes would tell us whether the outcome of the meeting was to tighten the rods.

Quote (epoxybot )

It raises the question whether they on site to perform work based on the smaller cracks and somehow the word wasn't circulated that a re-think was in the works.

The NTSB photos in the 9 Aug 2018 Update were dated 13 and 14 March 2018. It would be unimaginable the last big known cracks were not discussed at the 15 March meeting held five hours prior to collapse. Any engineer who has worked on site for say a few years and have seen strength tests by crashing concrete cylinders would know the bridge has gone just by looking at the cracking severity in the NTSB photos, one of which shows a tape can be inserted to 25% depth of Member 11 two days before the bridge failed!.

NTSB preliminary report shows the cracks before the bridge was relocated. The last known cracks should be those in the NTSB 9 Aug Update. Three photos were sourced from the Contractor MCM and one from BP&A Consultant most likely sometimes after the PT rod tension was removed in Member 11.

May be I am among the minority disturbed by the magnitude of the cracks. However the majority of the large cracks photos were taken by the workman who carried out work on the PT rods. My speculation is that he was equally surprised too and had gone out his way to record the evidence rather professionally, with a tape inserted as a comparison. He must have been deeply worried if he had done something wrong and the photos were his protection to show the end result of his work was as per instructions/commands. Sadly his employer still sent him out to do more work on the PT rods and he was seriously injured at the end.

Don't think this has been mentioned before, but the bridge may have been unknowningly damaged during the move. From the NBC Timeline:

March 10th:

March 15th:

Let's see how damage could have inflicted to the bridge during the move.

The FIU bridge has a dead weight of 950 ton. It has 5 bays so roughly each bay weighs about 190 ton. During the move the two end bays were cantilevered out so each might have a maximum of 190 ton hanging freely.

Member 11 has two PT rod each stressed to 280 kips so together the two rods offer about 250 ton resistance.
Since the bridge was tightly monitored and fully rigged with strain gauges I doubt much damage could have occurred.

The Member 11 has a slope of 35 degree (1 vertical:1.427 horizontal:diagonal 1.74) so if the reaction at the support were to be 475 ton then the axial compression in Member 11 due to the dead weight could have been about 950/2*1.74 = 826 ton

To me a real killer could be the omission or the lack of consideration of 250 ton from the PT rod stress, required only during the move, in rebar design of Member 11. 250 ton is an increase of 30% load to member 11.

FDOT had hand written the need of temporary PT rods on FIGG's drawing on 15 Sept 2016. Any engineer looking at the rebar inside Member 11 would know it could not possibly take 1/5 of the bridge dead weight. This raises the alarm about FIGG's competency have to rely on FDOT to tell them something so fundamental. If FIGG had implemented FDOT suggestion of the extra PT rods but did not follow through with the rest of the structure especially the connection design of Member 11 to take the extra 30% load then that might be the smoking gun everybody been looking for.

It was FDOT, that asked them to add room for an extra lane just before construction was due to start. I wouldn't have been real happy having to deal with that at the last minute. What I find mysterious is that members one and two were beefed up, but not eleven and twelve?

SF Charlie
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The strain gauge data may have been corrupted, allowing the bridge to be over-stressed without a warning indicator. BDI was responsible for installing and monitoring the strain gauges. They acknowledged to the bridge mover, Barnhart, that at one point a computer was turned off and on during a work break, and all their gauge readings were reset to zero instead of reloading the stored values that were present before the break. EDIT ADD: After re-reading the Timeline description, what may have been lost were the inital calibration offsets that were used to zero the gauge readings before the move began. This implies that many of the monitored areas could have exceded 100% of the safety threshold during the remainder of the move, and at least one of the monitored areas may have experienced close to 200%. It's also possible that the lack of concern about the cracks after the move may have been partly influenced by the "good data" provided by BDI showing the bridge hadn't been overly stressed during the move.

Multiple problems with the data collection process. The BDI software apparently wasn't configured to auto-resume after a reboot, or at least indicate a warning. The "meter reader" didn't notice the gauge readings were reset after the work break, possibly because a "new guy" took over. In either case BDI should have anticipated such a problem and trained their personnel. The data discontinuity at the break should have been clearly evident in the charts or readouts, and somebody from BDI or Barnhart should have noticed it during a post-move data review.

Given that there were only two load points on the bridge I don't think their issue was changes to the bending of the bridge and in any event the final two bays were hanging off the end so not really impacted by changes in elevation at the support points.

The single line of the columns but wide bottom flange gives rise to twisting of the bridge if the trailers encounter changes to the elevation which as they had to negotiate a verge and the central reservation is not surprising. Hence the limit of 1/2 degree of flex.

Remember - More details = better answers
Also: If you get a response it's polite to respond to it.

Quote (SFCharlie )

It was FDOT, that asked them to add room for an extra lane just before construction was due to start. I wouldn't have been real happy having to deal with that at the last minute.

From record the bridge was moved as one complete unit by 11' to the North. The span was not changed so structurally the structure was the same as before. As construction had not started yet so the impact to the cost should be minimal.

Quote (saikee119)

The span was not changed so structurally the structure was the same as before

That's the point: The span wasn't changed (enough) but the support during the move was changed fundamentally. Also the construction was scheduled to start within a month. All the adding the PT rods in 11 etc. happened after the support during the move was changed to accommodate the 11 feet of dirt adjacent to the center pier.

SF Charlie
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There was a non-trivial charge for the extra engineering for the 11 feet for the extra lane. Moving the center pier closer to the canal required driving extra piles taking more time and delaying the start of pouring the piers, putting the schedule in a time bind with the end of the money availability running out causing the rush at the end causing the panic which may have affected the critical thinking at the time of the move.

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MCM asked for an additional 180 days & $1.7M to effect the shift of the bridge.

The timeline was helpful to resolve my impressions that workers were up on the bridge while the meeting was ongoing. They were on the bridge at 10:40AM, with the meeting concluding at 11:00AM. What is concerning, is the anecdote that engineers were seen on the bridge, in the hour prior to the 9:00am meeting, at which Denny Pate gave a presentation and presumably stated; there were no safety concerns & the cracking did not compromise the structural safety of the bridge. Just who at the meeting was up on the bridge?

FDOT engineer Tom Andres' name has been dragged through this disaster by the media & by inference by FIU's finger pointing at FDOT but when he sent an email confirming all his concerns had been resolved, including the positioning of the SFMT transporters, the SMFT transporters were positioned differently under the 2-3 & 10-11 nodes. There isn't any source showing he was informed or approved the repositioning of the SMFT transporters further inboard from both ends of the bridge. No reason has been given for the changed positioning of the SMFT transporters. If he wasn't notified, I imagine having his name dragged through the press does not make him very happy. The timeline shows he doggedly pursued each of his original concerns to a conclusion.

Since we now know the bridge was UNDERNOURISHED, the real tragedy is that the telltale signs were present from the time the bridge was set to the time of its collapse and good people failed to act.

Does anyone know why the move was halted just before dawn? Stuart Grant took several time-lapse videos from different viewpoints as the bridge was moved, but this ground-level video, credited to Grant, no longer appears on his YT Videos page. It shows several instances of the bridge shifting on the north movers as the south movers were repositioned, then the move being stopped while the north mover is still partly on the center divider. After a short pause someone is lifted up to the center of the span, and the move continues immediately after he is brought down. This video shows the person making several trips to the south end of the span.

Quote (MikeW7)

Does anyone know why the move was halted just before dawn?

.

Most likely, the move was halted to await the arrival of dignitaries.

epoxybot - They paused while the bridge was it still partly on the center divider and before it was aligned North-South. If they were concerned about Show-N-Tell they would have waited until it was squared up with only a few feet to go. Looks like they encountered a problem, stopped to discuss it, then sent somebody up to have a look. If it was BDI checking their instrumentation, this may have been when their computer was shut off. Odd that they only checked the south end even though hairline cracks were previously found on both ends (members 3 and 10).

Here's a compilation video posted 2018 March 15 that includes extensive drone footage of the entire bridge move, Stuart Grant's ground-level time lapse of the move, and high-definition helicopter footage of the north end immediately after the bridge fell.

Looks like MCM Construction has settled with the vitim's families for $42 million. Link

From the local miami news:
Construction company reaches deal to pay up to $42M to victims of FIU bridge collapse

SF Charlie
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Maybe I'm not understanding the legal strategies here - but if the NTSB hasn't determined the cause of it - why, as a contractor, would you settle for $42 million if the cause might end up being totally on the designers?

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Quote (JAE)

Maybe I'm not understanding the legal strategies here - but if the NTSB hasn't determined the cause of it - why, as a contractor, would you settle for $42 million if the cause might end up being totally on the designers?

Yes, I agree, however, it is not like the contractor really has a role in accepting or rejecting a settlement. Once the insurance co's get hold of the case they and their legal team take it over, and the contractor just follows the lead. Bankruptcy of MGM may have also lead them to this settlement too.

This March 14 ENR.com article says MCM had been operating at a loss since 2016 but wanted to make sure they were able to pay out settlements, so they established a trust/fund to cover all claims as part of their bankruptcy reorganization. The $42 million figure was the maximum amount they estimated could be secured from their insurance policies.

EDIT ADD: Based on this May 03 Miami Herald article MCM reached an agreement with their insurers, not the families. My impression is that because of their ongoing bankruptcy problems, they wanted to make sure money is available in the future if any claims are eventually settled against them.

Quote (JAE)

Maybe I'm not understanding the legal strategies here - but if the NTSB hasn't determined the cause of it - why, as a contractor, would you settle for $42 million if the cause might end up being totally on the designers?

Not a lawyer, but based on what I've seen of past cases I think there's a couple things at work here:

* The NTSB report is pretty much a foregone conclusion at this point. Pretty much everyone who matters knows that it will demonstrate that the bridge was designed with inadequate shear margin to prevent what was essentially a tear-out failure of the 11/12 node. It will probably also demonstrate that it was built as designed, although that is unlikely to provide the builders much cover.

* Early settlement is in accord with two ancient axioms: "The bird in the hand is worth two in the bush," and "The first loss is always the smallest." Dragging this thing out is only going to make it uglier and more expensive to everybody on all sides (excepting perhaps the attorneys), and decrease the amount of money available to compensate survivors.

* Pure speculation, but similar past cases suggest that all involved in the design and construction and underwriting have gotten together, gotten their stories as straight as can be managed, and will present a united front that will limit further culpability to the degree practical.

I have the most scorn for those on site who saw the concrete fracturing off in huge chunks and did not declare an emergency, close the road and bail off that bridge. Shades of the Quebec bridge collapse where the on-site engineer kept piling up steel as compression members bowed. After that Wassisname who gave the order to proceed without going on-site and also not pulling everyone off the bridge and closing the road.

The contractor's insurance probably settled relatively quickly to avoid litigation. They had liability, so why prolong the proceedings with the attendant costs? As to the designers, I think they were working for the contractor, so that was not an available dodge for the contractor. Typically, construction litigation proceeds with a number of settlements before any remaining defendants have their days in court. There were earlier reports that the insurers of the designers were denying liability, so that will be more complicated to resolve. And then there will be a host of other defendants.

https://www.miamiherald.com/news/local/community/m...

More news, looking now for the 44 page document referenced in the article.

IC

Latest from AP.

• https://www.msn.com/en-us/news/us/records-cracks-in-miami-bridge-grew-daily

The 44-page document was released by FDOT:

https://cdn2.fdot.gov/fiu/14-FIGG-Structural-Analy...

Some interesting stuff. I wonder what exactly FIGG thinks they can dispute in the meeting minutes that look to match up with the presentation. I suppose they could try and take somebody along with them or possibly deny the presentation (initially) since the .pdf doesn't show their name on any of it.

This whole presentation shows a scary reliance on their modeling and calculations. I don't care what any of the math says, the bridge is screaming for help and they state there's no safety concern in the same presentation where they say it's not clear how the cracks formed. I'm going to double down on caution if a structure is behaving unexpectedly.

I didn't go through the full presentation but it didn't appear to me that they had any checks of the longitudinal reinforcement anchorage resisting the thrust from the diagonal. I got the distinct impression that they checked most everything but the most obvious limit state.

Ian Riley, PE, SE
Professional Engineer (ME, NH, VT, CT, MA, FL) Structural Engineer (IL, HI)

It's difficult to endure the presentation knowing that they completely lost the understanding of basic structural principles while ensconced in the defence of their position.

Shame on anyone that utters the term re-tensioning with regards to the work that was being done. What were they really trying to accomplish? Hang the bridge from itself to relieve the stress?

They were trying to use the PT bar to pull the broken sections back together without ever making a diagram that showed the PT line of action would do the opposite. Since the fracturing didn't happen until after the PT load was removed the naive choice was to put it back to stop the fracturing.

Such reliance on the LRFD and not a free-body diagram in sight.

Is there any surprise at the line in the chart "And therefore the is no safety concern ..." An obvious typo and sign of not paying attention to detail?

But they had fancy FEA screenshots
(/s)

Looking at pages 27 and 28, they clearly gave some thought to shear tear-out of the 11/12 node.

I don't know what went wrong, but I don't think I'm going too far out on a limb to make this guess based on the table on page 28: In assessing the shear resistance contributed by the rebar crossing the shear plane, they assumed that all members were loaded equally in shear. However, the evidence of the cracking shown in the photos suggests that the loading was not uniform; that members were sequentially overloaded and yielded, resulting in even greater loading of the next member along, and its subsequent failure, until the whole thing zippered off the end.

I wonder if FDOT engineer Tom Andres' notes were ever seen by the PEER review team.

They completely missed a potential failure plane in their presentation.

On page 30 a total shear surface area is shown of 23.62 SF. This whole presentation is made assuming shear failure will occur WITHIN the deck, thus the multiple of 2 being used on 11.81 SF. In this presentation they did not check a shear plane happening along surface of deck. Shear surface area in this instance would be roughly 1.75' (width) x 6.33' (length) = 11.1 SF. This is under half of what they've assumed. And, for this failure plane, there will be NO nodal confinement force from transverse PT, as failure plane is deck surface. I didn't run the numbers for the steel--no shop drawings, and I'm just too lazy at this point--but they're likely to be equally worrisome.

epoxybot, if you are referring to the suggestions that cracks could form, I think that was due to the observation that differential loading of the various portions from the PT bars would create too much strain in the tension sections in combination with differential shrinkage as the concrete cured.

I don't feel there was worry about immediate catastrophic failure because of the lack of adequate reinforcement but concern about exposure of internal reinforcement to corrosive elements.

Greenlama - that answers a question I would have asked - where did the multiplier come from. How did they not know about the cold-joint between the deck and the truss? For certain the photographs they had access to showed shear displacement along the deck and not damage within the deck beyond the spall-out on the end. It's like they determined at some time that the fracture at the deck end was because of transverse bending of the diaphragm, hence the concentration on the pads underneath.

I'd appreciate if someone savvy in structural engineering could clarify a few points in the recently-released"FIGG Structural Analysis Presentation Meeting Minutes" document. (https://cdn2.fdot.gov/fiu/14-FIGG-Structural-Analysis-Presentation-Meeting-Minutes.pdf).

1. The doc uses phrases like "capture the nodal zone" (referring to the 11-12 node). What sort of structure does "capture" imply adding?

2. My non-expert reading is that FIGG seemed more concerned with the vertical load causing problems (hence the shim recommendation) than constraining the horizontal load of #11. Is that a reasonable reading?

3. Page 27 gets to the topic of maintaining the connection of the 11-12 node to the deck, so as to be able to "capture the longitudinal force component of the strut"... ie: connect the strut's horizontal force to the deck's longitudinal PT bars. Page 28 shows some accounting of rebars crossing "assumed shear plane". Which plane would this be?

4. Page 29 assesses "nodal shear stability" in a way that might be problematic. If I understand correctly, this looks at transverse forces across the deck, east-west) compressing ("confining") the concrete in the region of the node. It assumes an even distribution of transverse tendons, spaced at 175/65 ft, or 2.7 feet, then applies the force implied to a 4.75-ft "side" of the region of interest. This 4.75 ft is presumably the length of the oblique connection of #11 to the deck.

However drawings such as this one: https://cdn2.fdot.gov/fiu/13-Denney-Pate-signed-and-sealed-FIU-bridge-construction-plans.pdf page 69 appear to show that there are no transverse tendons (or at most 1) in the end region of the deck or diaphragm, the first one coming at something like 3ft 8in from the end (on the plan view), or maybe 4ft 5in (View B-B), or maybe 4ft (page 71, view B-B) or maybe 4ft 2.5in (page 85).

So it seems to me that not only were there either zero or just one horizontal tendons in this area, but they were not distributed evenly over the area of interest, with the compression effect tapering off rapidly toward the endmost region of the deck where the longitudinal force (and its horizontal component) would be most focused. Certainly there will be some compression from the endmost transverse tendon, and diminishing amounts from the second, third etc, but the compression won't be in a strictly transverse direction.

Am I reading this right?

5. The FEA graphics on pages 36 through 43 I think treat the members as homogeneous objects, and led to the conclusion "the spalled areas have not been replicated by the engineering analysis" (page 44). As it turns out, I guess this is a reminder that if the analysis doesn't predict what is actually happening, then the analysis is wrong. Presumably one or more of the actual members were no longer functioning as homogeneous objects, but instead had internal cracks severe enough to require a more complex model for analysis.

The release by FDOT of this meeting record, presumably as recorded by the FIU attendees, is useful in showing the frame of mind of the participants, in particular the FIGG engineers. They were under pressure for sure, but had absolute faith in the correctness of their design. That faith was woefully misplaced, at the expense of a number of lives lost.

The NTSB is taking its time in doing this investigation. I just hope that they don't just focus on the immediate locus of where the truss/frame failed, but rather on the whole flawed concept.

gwideman, here are my thoughts.

1) "Capture the nodal zone" is an odd phrase. Took me a while to comprehend. Because this is a concrete truss, and thus unique, I think they were reaching for terms to help them in their discussions, and analysis. Maybe capture means how best to capture the behavior of the truss.

2) Yes.

3) Force here, best I can tell, is the axial force in diagonal. Shear plane is that shaded area on page 27. And there are two areas, one on either side of outside faces of diagonal extending into deck and diaphragm. And it should be noted here that with the slope of the diagonal, the horizontal component of diagonal force will be LARGER than the verticl component.

4) Once transverse PT is stressed and grouted the force cone will tend to evenly distribute across the deck, reaching a uniform distribution near the centerline of the bridge. You are right to question conditions at end of bridge. Here, transverse PT stresses would be expected to lessen. How much is open to discussion.

5) I think your take on the FEM is a good one. Concrete with large cracks should NOT be modeled homogeneously.

Quote (TheGreenLama)

Shear plane is that shaded area on page 27.

Ah, that makes sense, thanks.

And interestingly, that same figure shows the arrangement of transverse tendons, somewhat speaking to my question 4 about transverse forces containing the region where the axial forces of #11 are "captured". That endmost transverse tendon does not seem ideally placed to help, though it's not clear whether this drawing is to scale.

Quote (TheGreenLama)

You are right to question conditions at end of bridge

And on that topic, page 29 shows a calculation resulting in 520kips of confinement force. But that would be for a region of the bridge that has transverse tendons both north and south of it, and at 2.7ft spacing. Since the region of node 11-12 has only a tendon to the south, and only at the edge of this region of 3 or 4 feet length, I would guess that the actual confinement force resulting from the tendon would be reduced by a factor of at least three of four, so more like 130 to 170kips.

Following this to page 31, this would make the (Phi)(Vni) (without c) = (1908 + 1.4 x 130) x 0.9 = 1881
Which is less than the (page 32) factored Demand Nodal Shear of 1983.

This is a relative pessimistic version of the calculation, but it doesn't look encouraging.

Quote (1) "Capture the nodal zone" is an odd phrase. Took me a while to comprehend. Because this is a concrete truss, and thus unique, I think they were reaching for terms to help them in their discussions, and analysis. Maybe capture means how best to capture the behavior of the truss.)

Want to emphasize this isn't a truss; truss members only carry axial force and nodes do not carry moment. For this concept, the open web members have to carry moment, shear, and axial force, and the nodes have to restrain these forces.

And therein lies the rub. The design itself was very ambitious. How they would practically design a giant, simply-supported, all-concrete I-beam relying on frame action between the angled "web" members and the top and bottom flange is beyond me. I truly wonder if there was ever a sound way to pull this concept off. If they wanted concrete so badly, a solid web would have made more sense.

"...if the analysis doesn't predict what is actually happening, then the analysis is wrong."

BINGO! I think you hit on the crux of the failure to anticipate the imminent collapse of the bridge, gwideman.

The question now is this - will claiming total incompetence at understanding the mechanics of the failure with direct observation of the failure be enough for a manslaughter conviction for certain engineers at FIGG and the other companies or will it be a shrug that mistakes were made? How could they see the horizontal displacement at the bottom of 11 and the spalling of the faces and conclude the shims on the pylon were the problem?

I think the meeting notes hint that they knew what was really wrong and they were trying to cover for it when they indicated a desire to tie 11-12 back to the 9-10 node with steel channel until the backspan and pylon were finished; the "capture" mention in the notes.

Questions & Answers
CEI to FIGG: Do we need temporary shoring?
o FIGG responded that it was not necessary. Rather than carry weight, carry load off that number/node. Steel channels to 10/9 node & PT Bars to capture some of that force which is better than vertical support. The diagonal member is what needs to be captured

I think they believed the reapplication of post tension would pull the end of 11 back somehow and that the steel channels would be better than putting cribbing under the bridge as cribbing would make the failure plain to everyone.

Quote (3DDave)

tie 11-12 back to the 9-10 node

Ah, I had seen "Steel channels to 10/9 node & PT Bars to capture some of that force which is better than vertical support.", but had misunderstood that this was addressing some separate issue they had with 10/9. Your answer filled in that this means channels/PT bars FROM 11/12 TO 10/9. So something like this sketch, which also includes the proposed shim in the gap left directly under 11/12:



And "which is better than vertical support", because just supporting near the end of bridge, doesn't deal with the fact that, as a truss, the top member is in compression, which needs to be resisted by the horizontal component of compression of 11. So plausible vertical support would need to have been inboard, say under 9/10 like when the bridge was moved. And that would block traffic, and might require yet more calculations unless the other support used during moving was also put in place.

Quote:

Want to emphasize this isn't a truss; truss members only carry axial force and nodes do not carry moment.

Tell that to a Vierendeel truss.

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Many of the comments here deal only with axial forces, which mirrors the problem with the design concept. As samwise753 described above, this was a concrete frame. The Vierendeel action of the frame was not adequately considered.

But it is worse than that...the internal restraint stresses were apparently not well thought out either.

Looks like I was typing at the same time as JAE was posting.

About "comments here deal only with axial forces" and "Vierendeel" action, and "this isn't a truss".

Clarification please: Are you guys saying that the FIU bridge structure is being discussed here (and perhaps designed) as though it was a truss, but in fact lacks the compliance at joints to be considered as such; that the concrete joints will attempt to resist rotations (as in Vierendeel designs), but not being appropriate to do so will be damaged under certain conditions?

Quote (hoki66)

But it is worse than that...the internal restraint stresses were apparently not well thought out either.

At some level the 11/12 failure appears to be insufficient ability to transmit axial forces to the deck's longitudinal PT bars. Those tendons are positioned in a way that they won't do much to support the deck by itself, so their main objective is to resist the horizontal tension "of a truss's lower members", a substantial amount of it imparted by the two end diagonal members.

Evidently 11 was supposed to transmit the horizontal component of its axial force across the end diaphragm to the PT bars (and the vertical component across the diaphragm to the support pads which were not directly under 11/12).

Has any documentation been released on diaphragm structure, and how 11/12 was connected to it? To what extent does it rely on the integrity of the concrete (possibly damaged by one or another effect, such as rotation of the joint during transportation), as opposed to the capability of the rebar unaided?

(And I note that this ties in to my comment a couple of posts ago that the confinement force for this area seems to have been calculated based on an unrealistic contribution from the transverse tendons, at least for the March 15 2018 meeting.)

Not the way I would have said it, but yes, this bridge was a frame, and the members and joints needed to take a lot of bending moment, which I don't believe was adequately considered. Rigid joints in concrete are not easily achieved, in fact never achieved 100%. And the other issue I have raised numerous times was about the internal restraints. Understand that problem?

Yes, agree with hokie66.

The "truss" as it was, had all diagonal members and thus - stable triangular shapes along its length.

For most historical truss design (with steel) engineers traditionally assumed that there would be a bit of give in the diagonal end connections, or that if you ignored the fixity at the diagonal ends, there would be some slight inelastic response that would allow you to assume the diagonal ends were moment free (pinned). This also simplified analysis prior to our handy PC's.

However, with this "truss", there was true fixity and the ability of the diagonal ends to be axially connected to the decks depended on the integrity of the diagonal-to-deck connection, which appeared to have been significantly compromised by non-uniform shears though the deck....due to moment perhaps.

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This structure was unique, in the history of the world. Sure, it was built with concrete and steel reinforcement, but the conceptual form was unique. And there was no testing of the concept. No matter how clever or innovative it was thought to be, it was illogical to jump headfirst into this concept without any physical testing. In my previously expressed opinion, that testing should have been load testing of the actual frame while it was still at the casting location.

Touche, JAE. I didn't remember a Vierendeel Truss, though it is a truss in name only.

Well said, hokie66. Rigid joints with members at odd angles was just crazy.

From the presentation, they relied on shear friction to tie the members together. Not sure if their computations or the actual details comply with accepted practice, but if so, then the whole concept of shear friction would need reexamination. We have had lots of arguments in the structural forums about shear friction theory, and I for one have not accepted its validity.

Quote (gwideman)

At some level the 11/12 failure appears to be insufficient ability to transmit axial forces to the deck's longitudinal PT bars. Those tendons are positioned in a way that they won't do much to support the deck by itself, so their main objective is to resist the horizontal tension "of a truss's lower members", a substantial amount of it imparted by the two end diagonal members.

I would say, "all of it imparted by the two end diagonal members."

Here's a thought experiment. Look at member 11 as a column framing into a slab. If member 11 was vertical we would check the "punching shear" capacity of the slab for the axial load in the column, similar to what was done in the recent FIGG presentation. And assuming no moments were in the column, only nominal longitudinal steel would be required. But once the column enters the slab at an angle, as is our case, longitudinal steel from member 11 is going to be required to get the axial load into the slab (and thus the longitudinal PT). That's what's so puzzling about the FIGG presentation. Why didn't they also run numbers for the horizontal component of member 11? They've also failed to detail hairpins tying member 12 back into the slab. So is this a case of people at FIGG not seeing beyond a FEM model?

Quote (hokie66)

this bridge was a frame

Ahem, just now catching on that "frame" has a technical meaning that distinguishes it from "truss", in that frames have joints that can transfer moments (resist rotation), whereas in a truss the members are considered to be pinned, and this able to rotate (at least a little) at the joints, conveying only axial forces.

So if I understand the discussion properly, the idea is that the FIU bridge could work fine as a truss, say made of steel, where some (small) amount of rotation at the joints would be acceptable. Or with steel some or all joints could be made rigid, and achieve some joint-bending-resistance frame properties. But implemented in concrete+rebar, significant trussly bending at the joints would cause damage, and framesque resistance to bending would be difficult to implement.

Quote (hokie66)

And the other issue I have raised numerous times was about the internal restraints. Understand that problem?

By "internal restraints" are you referring to (competing) forces/moments operating within the concrete in the 11/12/diaphragm region that could cause internal cracking, for example during concrete curing and shrinkage, or loading?

And that would shed light on whether or how the concrete would maintain its integrity and be able to perform, with the rebar, as simple homogenous components?

Are there docs which bear on this?

The pessimist in me says they were clearly understanding the horizontal shear failure and that's why they wanted to tie back to 9/10 with the steel channels until they could cast in place a stop in the form of the other span and the pylon.

FIGG's is the presentation I would make if I wanted to point the audience in a different direction from the actual failure; however it seems like anyone who did the FEA should have been smart enough to realize the fundamental flaw and not provide anything else that could be used to distract.

However if I did understand the true failure mode the last thing I would authorize is retensioning the PT bars in 11. So I don't know how Pate got himself to that presentation, except brutal ignorance. He knows 11 and 12 are sliding out, has a plan to stop them from sliding out, yet has no grasp on how the are sliding out. And, apparently, neither does anyone else in the room or about to apply the hydraulic jacks to the PT bars.

I think the critical question that might have worked is - why aren't they sliding any faster? That brings focus to the re-bar and from there the loads that have to be acting on the re-bar, which should produce a free-body diagram showing that the diagonals are not actually tied to the deck except by the re-bar; which the PT bars cross. If they had put extensiometers across the cracks in the direction of the movement they would have been able to get a feel for if the failure was speeding up or slowing down and therefore if the re-bar or re-bar anchorage was failing or the re-bar was aligning itself to better resist the load.

In the meantime someone would certainly have called for at least one crawler to come back and shore up the bridge since all it would take for a cascade catastrophic failure is one strand of re-bar to give and transfer it's load to all the others, zippering though them in quick order.

gwideman,
That's the gist of it. The various members of the bridge wanted to change volume at different rates and at different times. Therefore, shortening due to shrinkage and prestress, as well as perhaps temperature, would have meant that the web members were acting as struts to prevent the shortening of the decks. Restrained shortening places concrete in tension, and it cracks. But this structure was never "homogeneous", as it was cast bit by bit.

I have atttempted to recreate the "total nodal shear stability" accounting on page 28 (and following) of the FIGG presentation https://cdn2.fdot.gov/fiu/14-FIGG-Structural-Analy....

The table lists "Rebar crossing assumed shear plane", where the "assumed shear plane" is that which is important in conveying the axial force of member 11 to the deck's longitudinal tendons.

I have cross-referenced the listed bars to the rebar list and detail drawings in https://cdn2.fdot.gov/fiu/13-Denney-Pate-signed-an....

Some of these rebars don't appear to me to cross the shear plane.

From page 28:


Correlated to the rebar list, and color coded by diameter:


My attempt to find them on the construction drawings. For a larger version see the attachments. Likely you'll need to look at the PDF to see the shape and extent of the rebars in detail.


And here are my comments on each rebar line:


In short, of the 22.72 in^2 called out in the presentation, I only found 9.99 in^2. Have I missed something here? Perhaps these were not the final construction drawings?

Now considering the calculation on page 31:


...I commented before that the Mu x Pc component looks overestimated by a factor of three or four due to the geometry relative to the endmost transverse tendon. Now the accounting of the rebars crossing the shear plane(s) looks to be about only 40% of what the presentation counts.

That line of thinking would lead to total Vni being mostly dependent on the Cohesion portion, and without that, significantly less than the factored Demand Nodal Shear of 1983 kips (page 32).

And here's the attachment.
https://files.engineering.com/getfile.aspx?folder=...

• https://files.engineering.com/getfile.aspx?folder=9aa25502-8318-466f-8b32-f

Quote (gwideman)

Perhaps these were not the final construction drawings?

Even though these were 100% sealed, there must have been later construction drawings, because these do not show the span shifted to the north, with the pylon base in the SFWMD canal. I am not an engineer but rather know waters well, and I've always wondered about how the shift affected the stability of the north end, considering that a new retaining wall and fill were needed; previously spec'ed concrete piles were now used in the canal instead of on the bank; and whether dredging in that canal location (which was occurring during construction) had any effect. The shift was dictated by FDOT in October, so it's weird that plans signed in December still show the pylon on dry land.

Oh sorry, those were the April plans, not the December ones. Maybe they did show the shift?

Gwideman, For the transverse PT, they're not talking about how many PT tendons run thru the diaphragm. They're talking about the stress the transverse PT generates along the centerline. And for this you need to consider all the PT. Say you conservatively assume a stress cone extending out at 45 deg. from each anchorage, and the spacing of anchors is 2' - 8". That means that only in the outer 2' - 8" of the deck is there a reduced stress. Looking at the drawings again, the first transverse PT comes in 4' - 0" from the end, and then the 2' - 8" spacing begins (Dwg B-60). Based on this, I think I'd like to modify what I told you earlier about forces lessening near the end. I think that for the purposes of what FIGG is attempting to demonstrate in that calc, using a 54.8 k/ft as an approximation of the transverse tendon force along the centerline is OK.

As far as whether these are the final drawings, the answer is no. There are shop drawings, which are prepared by the mild reinforcement provider, and reviewed by the designer, which detail exactly size and placement of steel. Since these are created based on the design drawings, they're generally identical. But we don't have these. Then, once the bridge is finished, an as-built set of drawings is usually created, which includes any modifications that were made to the original design plans during construction. We don't fully know what these are either.

Lacking shop drawings is one reason I refrained from delving into that very crowded area of the diaphragm. Not sure I want to try and tackle it without them.

Quote (3DDave)

FIGG's is the presentation I would make if I wanted to point the audience in a different direction from the actual failure; however it seems like anyone who did the FEA should have been smart enough to realize the fundamental flaw and not provide anything else that could be used to distract.

Misdirection--that's something I hadn't considered. FIU, the owner, was an attendee. 3DDave, you may be on to something.

GreenLama - I want to clarify that it's very likely that Pate & Co initially mislead themselves, the presentation was a continuation of that process; before this presentation they'd already set the PT crew do exactly what would increase the forces in the direction of failure.

I said a similar thing for the failure of MCAS - it all started with a convincing presentation. This follows that pattern.

Gawd how I hate PowerPoint. It's like people switch off their brains when they make them and when they watch them. Just because it's on a screen does not mean it's true. Make a real report, unroll actual blue-prints. Put an actual pencil to actual paper. Convene at the failure site.

PowerPoint in the news recently:

The PowerPoint slide that crashed a Space Shuttle
General Mattis, save the U.S. military. Ban PowerPoint.

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Quote (TheGreenLama)

For the transverse PT, they're not talking about how many PT tendons run thru the diaphragm. They're talking about the stress the transverse PT generates along the centerline.

Yes, exactly, that's what I was talking about with relation to the Mu x Pc component of the overall page 31 calc.

I don't know the theory that governs how the transverse tendon tension translates into a distribution of compression at the (longitudinal) midline of the bridge. However, my naive assumption is that away from the ends of the span, each point at the midline receives a sum of forces from the transverse tendons to its north and south, most from the closest tendons, and diminishing force from those further away. The further away the tendon, the more those forces are not perpendicular to midline, but the north-south components of the numerous contributions will cancel.

So it's as though each 2ft-8in section of bridge is compressed by one tendon; each tendon serving 1ft-4in region to its south and to its north. Construction drawings show tendon force after tensioning of about 160 kips, distributed over 2.7ft = 59 kips/ft, pretty good agreement with the presentation's value of 54.8.

But the end of the deck is only adjacent to a tendon to its south, with obviously no tendons to its north. So, again naively, that last region of 4ft is covered primarily by "one side" of the last tendon, and no tendons to the north. So the force that the tendon would normally apply to 1ft 4in of midline is instead distributed to 4ft, resulting in only one third the compressive force per unit length.

Clearly this is simplistic. No doubt the material complies somewhat to redistribute the forces a bit more evenly. But I am fairly convinced that the end region can't be seeing the same compressive force (from the transverse tendons) as a region in the middle of the span.

Since the failure did not happen within the material of the deck and there is not a good way to generate 1.5 Million pounds of resistance with transverse containment in this configuration, it's a moot point as to what the effect is.

Quote (3DDave)

it's a moot point as to what the effect is

I infer you're responding to my post, inquiring into the effect the transverse tendons would have in the region of 11/12?
It may be moot if one's position is that this was a flawed design from basic principles, no need to analyze the details to get to that conclusion. Fair enough.

However, it's not moot in trying to understand FIGG's reasoning. As you pointed out "it's very likely that Pate & Co initially mislead themselves". I think it is valuable to understand how engineers (or management, or whoever) gets into that position. In this case, were these qualitative mistakes (wrong type of structure for the job), or errors of magnitude (should have been 50% beefier), or of construction (drawing said X, but Y got built) and so on.

The recently-released presentation gives a little insight into the model they were using to analyze the 11-12 connection, and the numbers they were plugging into it. Was that a sensible model? Did those numbers properly apply to what was built (or at least drawn)?

Quote (3DDave)

the failure did not happen within the material of the deck

I'm not quite sure what you mean here. I think one of the prominent theories in this forum is that the failure did involved the intersection of deck and 11/12, with pictures before the collapse showing significant spalling of the deck surface adjacent to 11/12. Granted that was not the only spalling, and this is not the only theory, of course.

Third post in a series looking at the FIGG presentation Total Nodal Shear Stability calculations, page 23. This time, the Cohesion contribution.

The cohesion contribution is based on a shear plane area of 23.62 sqft, which is the shaded area on page 27. Here's how that number is arrived at:



, which agrees with the presentation (23.6).

However, that shear area is interrupted by two sleeves (around vertical reinforcing bars). These are shown in this figure:



The sleeves are each 4 inches in diameter, and have an intervening gap of 2.75 inches (lighter pink), which I think would not contribute to any shear force. So I suggest that the entire 0.9ft x 4.3 ft should be discounted from the shear plane area. (I have also marked a gray "???" area that I think also would not provide useful shear force, but I did not deduct it from the area).

Revised shear plane area: 23.6 - (2 x 4.3 x 0.9) = 23.6 - 7.7 = 15.9 sq ft.
Revised as percent of presentation = 15.9/23.6 = 67%

For orientation, here is the area under discussion. The sleeves marked with yellow arrows in the top picture are the ones also shown in the bottom picture. The sleeves marked with red arrows are on the other side of 12, and protruded from the deck only a couple of inches (as other pictures elsewhere in this thread show).



Shearing (or at least lack of cohesion) looks to have occurred around and between the sleeves. (Though this doesn't necessarily imply that this was the very initial part of the failure.)

The intersection, not the embedment. Since 11/12 weren't actually captured in the deck pour they could not be retained by the deck no matter what. Only the re-bar crossing that boundary could be captured and it looks like the portion in the deck remained there when the rest was sheared off. There's no reason to believe that the few bits of re-bar at the deck edge which carried the loads that spalled the deck could have been reinforced enough to make a difference considering that there were about 26 sections of re-bar joining the deck to the truss at that location; sections that all failed.

I'm glad new information is available now. Hopefully soon we will have a mathematical explanation for the failure.

I found reviewing MCM original technical proposal interesting > Link (Link typo corrected May 12 2019)

FIU bridge project homepage > Link

FIU documents homepage, includes calculations > http://facilities.fiu.edu/projects/BT-904-PRR.htm

Summary of series of posts examining the FIGG presentation calculations around node 11/12/deck "Total Nodal Shear Stability".



The equation has three contributions, each of which I've discussed in a previous post, and noticed that there are reasons to believe that each one is overstated in the presentation:

1. Cohesion component: Presentation uses 23.6 sqft as the area of the shear planes (total of the two sides), but that area is interrupted by sleeves that reduce this to 15.6.
Presentation overstates by 50%.

2. Rebar shear component. Presentation uses 22.7 sqin for the rebar crossing the shear planes (total two sides). I was able to find only about 10 sqin of the listed rebar (ie: only about 45% of it). That would put the presentation overstatement at over 120%. However, it's possible that the large bars that I either couldn't find, or that appeared not to cross the shear plane, were perhaps detailed in on some other drawing. (It might be feasible to correlate the drawings with post-failure photos.)

3. "Clamping" component: The deck's transverse tendons squeeze on the 11/12/deck node clamping it in place. The force used in the presentation is the average effect that the transverse tendons would have away from the ends of the span. My critique is that there is a uniquely large gap from the end of the deck to the first tendon, and that the area in question is only subject to the tendons on its south, with none on its north, unlike the middle of the deck. As a consequence, I (admittedly naively) think that the presentation overstates the clamping force by a factor of two or three. (Ie: Actual clamping force is only 33% to 50% of the presentation value.)

Summary of the Total Nodal Shear Stability components:


If the observations I've made here are valid, then obviously it's disappointing that such errors were made for the presentation to the crucial meeting, especially if they reflect the calculations done in designing the bridge in the first place, regardless of whether they turn out to be the root of the failure.

However, I reiterate that this is not my field of expertise, so certainly should not be taken as gospel.

Quote (3DDave)

The intersection, not the embedment.

I understand your points. I've not yet seen a photo from an angle that completely convinces me that 11/12 sheared off at the level of the top of the deck, but not an expert.

Regardless, I thought it useful to try to follow the reasoning and calculations in the presentation, because even if they were looking at the wrong place (vertical shear planes beside the projection of 11 into the slab, instead of the horizontal 11-12 to slab interface), it shows what they were and weren't taking into account.

And thanks to jrs_87's post of links, here we have some nodal zone calculations for the beefier 1/2/deck connection.

This pdf doesn't include any calculations for the 11/12/deck connection.

http://facilities.fiu.edu/projects/BT-904-PRR.htm > Calculations (zip).
> UCPP_Final_Calculations_Superstructure (1).pdf
PDF page number 1299 and following.



... and a couple more pages.

And there are also pertinent calculations and design notes in document UCPP_Final_Calculations_Superstructure Misc Details.pdf
Section I Deck End Diaphragms.
"A strut and tie model was developed for the Type I and Type IV diaphragms to determine the steel area required for the tension tie at the bottom of the section between the two bearings. The tension tie, compressive strut, and node regions were designed per AASHTO LRFD 5.6.3.4, 5.6.3.3, and 5.6.3.5 respectively. Crack control reinforcement and shrinkage and temperature reinforcement were provided in accordance with AASHTO LRFD 5.6.3.6 and 5.10.8 respectively. The bearing replacement case was also checked, but found not to govern the design. Shear friction at the interface of the diaphragm and typical deck section was checked per AASHTO LRFD 5.8.4.1. A similar analysis was performed for the Type II diaphragm during casting (section supported by bearings) using construction loads. The results of this analysis were conservatively applied to the Type III diaphragm design. [...]"

Quote (gwideman)

But the end of the deck is only adjacent to a tendon to its south, with obviously no tendons to its north. So, again naively, that last region of 4ft is covered primarily by "one side" of the last tendon, and no tendons to the north. So the force that the tendon would normally apply to 1ft 4in of midline is instead distributed to 4ft, resulting in only one third the compressive force per unit length.

Yes, but the tendons near the end of the bridge are acting over a smaller area of deck, thus the generated stresses increase. In other words, a tendon at the centerline of the diaphragm would, if it existed, generate twice the stress than one at midspan.

Gwideman, just as a matter of comparison, if you use the FIGG calc you've shown as a guide to calc the area for node 11/12, and not the hashed area shown in the presentation, the approximate shear area would be:

6.33' L x 2' D = 12.66 SF x 2 sides = 25.32 SF

In this case the conduit area would account for a smaller percentage of total area. Still, I think it would be correct to discount that area.

This is the first I've heard of the calculations being available online. I will have to look at when I get time.

And just a thought on rebar disparities between drawings and presentation. It's possible the shop drawings use different call numbers to identify the bars, and those call numbers are what they're showing in the presentation.

Quote (TheGreenLama)

In other words, a tendon at the centerline of the diaphragm would, if it existed, generate twice the stress than one at midspan.

Are you sure about that? If what you say is true, that seems to make the total stress higher than the total force applied by all the transverse tendons.

If there were N tendons, equally spaced X distance apart (with the end ones at X/2 from each end), each applying force F, then the force applied would be NxF, and according to your discussion, the resulting stress would be (N+2) x F.

Or have I fallen into some reasoning trap?

Gwideman, Assume distance from edge of deck to centerline is 10', and force from tendons distribute out at 45 deg. Tendon in center of bridge would have force distributed over 20' of deck. Tendon at end of bridge is only distributed over 10' of deck. Assuming same tendon force, stress would double.

Quote (TheGreenLama)

Assume distance from edge of deck to centerline is 10', and force from tendons distribute out at 45 deg. Tendon in center of bridge would have force distributed over 20' of deck. Tendon at end of bridge is only distributed over 10' of deck. Assuming same tendon force, stress would double.

I believe your point is illustrated like this:


... where this is a considerable simplification in that the forces won't distribute evenly over the entire +/- 45 degrees, and won't stop abruptly at the edge. However, OK for this illustration.

I understand your point, however, in the middle (longitundinally) of the span the midline points are receiving forces from PT bars within +/- 45 degrees, whereas at the end of the span they only receive forces from half that amount of PT bars [*], cancelling out the doubling that you state. Like this:



[* or less in the FIU bridge, due to the non-uniform wider spacing at the end of the deck].

I still might not understand something about this, but this is the way it seems to me.

[Edit: A few posts down, IceNine corrects my mistaken impression. The calculations I reference here are not for the endmost strut nodes per se, they are for the end diaphragms transmitting vertical force from those nodes to the support pads under the diaphragms. See later discussion.]

I'm sure everyone is enjoying their Saturday studying the Plans and Calculations docs at http://facilities.fiu.edu/projects/BT-904-PRR.htm

Pursuing further the design of the 11-12-deck area, document UCPP_Final_Calculations_Superstructure Misc Details.pdf presents some calculations.

However, despite this being a "Final" document, the numbers used seem disturbingly at odds with the actual bridge.

End Diaphragm I, PDF page 11

Struts 1 and 2, South end, not the ones that failed
Fc = 1336 kip
alpha s (theta s in the accompanying drawing) = 39 deg, compared to actual angle of strut 2 = 23.9 deg.

End Diaphragm II PDF page 31
Strut 11, North end (the one involved in failure).
Diaphragms II and III are considered together, despite dissimilarities.

Fc = 729 kip (this is half of what's reckoned elsewhere, for example in "the FIGG presentation" PowerPoint.
alpha s (theta s in the accompanying drawing) = 53 deg, compared to actual angle of strut 11 = about 32 deg.

Of course, the actual angles being much smaller in both cases than the ones in the calculations cause the actual forces to be considerably larger than those in the calculations.

There must surely be some later calculations, since these are dated September 2016, and don't seem applicable to the bridge as built.

gwideman,

These are strut and tie models of the end diaphragms. Because the centerline of the truss isn't supported with a bearing directly beneath it, the end diaphragm must span between bearings. So the angles in these calculations are the the angle from the centerline of truss to the bearing, not the angle from the truss member to the deck.

Quote (gwideman)

this is a considerable simplification in that the forces won't distribute evenly over the entire +/- 45 degrees, and won't stop abruptly at the edge

The distribution does stop abruptly at end of bridge. That's the whole point. Less deck area to capture force means higher stresses.

Quote (gwideman)

whereas at the end of the span they only receive forces from half that amount of PT bars

Yes, but the stress from those bars is higher than at midspan.

Quote (IceNine)

These are strut and tie models of the end diaphragms. Because the centerline of the truss isn't supported with a bearing directly beneath it, the end diaphragm must span between bearings. So the angles in these calculations are the the angle from the centerline of truss to the bearing, not the angle from the truss member to the deck.

Ah, thanks for that comment, I see you're right. So the calculations in UCPP_Final_Calculations_Superstructure Misc Details.pdf on PDF page 31 and following correspond to the meeting presentation slides thus:



... and the results compare like this:


... with the presentation's unfactored 593 (652 factored) considerably higher than the calculations' value of 439 which appears to be the factored value. This is I guess due to changing from one pad on each side of the support close to center line to two pads each side, somewhat further away, especially the outboard pads, hence more acute angles.

Luckily, the the eight #11 bars anticipated in the calculations (8 x 11S01, drawing B-47) are enough for the load. Though the presentation calculation adds in an additional 2 x 0.31 sq in for a slightly improved result.

Anyhow, this does not seem to be an element that failed, but it does mean I'm still not seeing where in the calculations the connection of #11/12/deck was analyzed.

Quote (TheGreeLama)

Quote (gwideman)

this is a considerable simplification in that the forces won't distribute evenly over the entire +/- 45 degrees, and won't stop abruptly at the edge

The distribution does stop abruptly at end of bridge. That's the whole point. Less deck area to capture force means higher stresses.

Perhaps the sentence you quoted was not quite clear. I am saying that for this discussion I accept your simple model of the PT force being distributed over a +/- 45 degree triangle, but I note that the nature of the simplification is that at the bridge midline the force won't actually distribute evenly over the +/-45 degrees, and also won't stop abruptly at the edge of the 45 degrees.

Yes, agreed, the forces will not exit the edge of the concrete, as I diagrammed right before the part you quoted.

Anyhow, the more important part is this:

Quote (TheGreeLama)

Quote (gwideman)

whereas at the end of the span they only receive forces from half that amount of PT bars

Yes, but the stress from those bars is higher than at midspan.

So at this point I think you agree that the two effects I diagrammed are indeed at work. And hopefully that in an "ideal" case where the transverse PT bars are evenly spaced, (so a half-space at the end), the two effects cancel out, resulting in uniform distribution of force to the midline of the bridge, and not an increased distribution of force toward the ends.

The one step further that I take it is that the actual bridge is "missing" the final tendon, so the transverse force at the midline near the end (at the 11/12/deck node) must be reduced.

I found a hint about shear with the deck on page 1382 of the calculations report. It refers to "Traction Force location from F.E." which doesn't seem to be mentioned anywhere else. There's something promising on page 1388, but it doesn't track with any force values I've run across.

Quote (3DDave)

page 1382 ... page 1388 ... doesn't track with any force values I've run across

I think you are on to something there. I'll try to elaborate.

As 3DDave has noted, pages 1382 through 1388 present some calculations that appear to calculate the forces at a horizontal shear plane that more or less matches the leading proposal for that actual failure location. But the numbers in those calculations don't match very well what we think we know about the bridge.

To try and correlate them with other numbers, I will look first at the #1/#2/deck connection, because those calcs almost correlate. Then look at the #11/#12/deck calcs.

First, here's what the forces should look like at 1/2/deck, I think.


These are based on assuming that there's a vertical component of 950 kips, half the weight of the bridge, through each of the end diagonal members. This matches the approach in the "presentation" PowerPoint.

Here's the shear plane referenced in the calcs:

(Yellow is in the pdf. Red and blue added by me.)

And the analysis results.

(Yellow is in the pdf. Red and blue added by me.)

No explicit angle is provided. However, I have added a sketch of the vectors from which to infer an angle.
The angle doesn't match what we might expect. But at least the vertical component is more than half the weight of the bridge. Possibly coincidentally, the horizontal component almost matches half the weight of the bridge. But is far less than the horizontal force one would expect from the angle of ~24 deg

A further oddity is the large Mxx value, which I assume is moment around the X axis. This perhaps corresponds to the offset between the center of the shear patch and the supporting column? (The small blue and red squares in the previous figure). 2.6ft x 1235 kips = 3211 ft-kips, very close to Mxx = 3113.

Comments: Very odd that the angle is so unexpectedly tall, but at least it has a vertical component that is in the same range as the half-weight of the bridge. (Though the concern of the analysis is to assess the horizontal component, that looks very understated.) I don't fully understand the analysis, but the force numbers are somewhere near the expected ballpark, if not actually in it.

For additional comparison, here's an excerpt from the analysis on page 1299:


Here again, puzzlingly, the vertical component is significantly less than the half-weight of the bridge.

[Part 2] Now for the 11/12/deck analysis from UCPP_Final_Calculations_Superstructure (1).pdf page 1388.

Here's the analysis diagram of the shear plane considered, and the forces involved, based on the assumption that member 11 has to include the vertical force of half the weight of the bridge. (Notably, that assumption accords with page 27 of the FIGG presentation PDF.)


(Yellow is in the pdf. Red and blue added by me.)

Here is the analysis result. Again I have added a vector diagram from which to infer an angle.

(Yellow is in the pdf. Red and blue added by me.)

The obvious issues are:
-- The analysis vertical force of 646.7 kips is substantially less than the 950 kips half-weight of the bridge. But oddly similar to the vertical force of 673.6 in the hand-written page 1299 notes.

-- The angle of about 48 degree is far steeper than the actual angle of about 32 degrees.

-- The horizontal force of 570.9 kips is far less than the expected 1531 kips.

These numbers are so extraordinarily different that I would expect someone to jump in and say that I've completely misunderstood the analysis, if it weren't for the results of the #1/#2/deck analysis, which, while still using an unexpected angle, at least had somewhat recognizable sizes of numbers. So I don't know what to conclude.

The other puzzle is that there are several areas where the number and spacing stirrups was done to keep the diagonals from rupturing, yet nothing about the amount of re-bar to take the shear loads at these locations.

I would like to make a friendly suggestion that any posters that markup calculation drawings and sketches come up with a method to distinguish original and contributed elements. I make this point only because the originals are by nature marked-up already.

Quote (3DDave)

The other puzzle is that there are several areas where the number and spacing stirrups was done to keep the diagonals from rupturing, yet nothing about the amount of re-bar to take the shear loads at these locations.

Agreed! There are such calculations of forces and comparison to the quantity of rebar in the calculations for the end diaphragms. Maybe there are some somewhere for the shear plane we're talking about here.

Also, I edited my 1/2/deck post to add a calculation that appears to confirm that the Mxx=3113 is moment attributable to the offset of the center of the shear patch from the support column.

If that's a correct understanding, then the structure would also need to withstand that moment, with appropriate calcs for the concrete and steel. Or the effective shear patch would be considerably smaller and more towards the end of the deck.

A bit of a sidebar at this point though, since it's 11/12/deck that's of most interest.

Quote (jrs_87)

I would like to make a friendly suggestion that any posters that markup calculation drawings and sketches come up with a method to distinguish original and contributed elements. I make this point only because the originals are by nature marked-up already.

Agreed, and good point. For my part I try to add any markup in a distinctive color, usually red or blue. Of course, readers may not be aware that these are my markups, but I also try to reference the original docs, which are readily available, and where the contrast will be obvious.

Quote (3DDave)

"Traction Force location from F.E."

The horizontal locations of the south and north edges of each of the shear plane patches are annotated in terms of "Global Z" from the south end of the bridge.

Perhaps the comment you quoted is saying that those Z values came from the F.E. model?

I was hoping that if they went to the trouble of mentioning where the force was located that they would mention what the force was.

FDOT documents page, updated May 6, 2019 to allow access to files previously blocked by NTSB.

https://www.fdot.gov/info/co/record.shtm

May 16, 2019 ENR article by Scott Judy and Richard Korman

https://www.enr.com/articles/46867-spreading-crack...

In the May 16, 2019 ENR article linked by jrs_87 it has this quotation from Bill Gamble:

Quote (Bill Gamble)

William L. Gamble, professor emeritus of civil and environmental engineering at the University of Illinois in Urbana, told ENR that he was “dumbfounded, greatly surprised and appalled” at the documents detailing the meeting, adding: “Yes, reinforced concrete structures often crack, but no, the cracks should not be large and growing day by day.

“I still find it hard to believe that anyone who had taken, and passed, a course on reinforced concrete design and behavior could not be greatly concerned,” he added. “Cracks that one can stick the end of a tape measure into are a collapse waiting to happen.”

Bold emphasis added by me.

[Revised to add image of spreadsheet calcs, which correct a miscalculation in previous version. Basic conclusion remains the same].

OK structural people, could you take a look at this and tell us whether I've understood the following?

If I'm right, the crucial page of calculations that verifies sufficient strength of the strut-to-deck nodes seems to have some conspicuous problems.

Quote (3DDave)

The other puzzle is that there are several areas where the number and spacing stirrups was done to keep the diagonals from rupturing, yet nothing about the amount of re-bar to take the shear loads at these locations.

I and others had been puzzled by not finding the calculations where the assessments of horizontal shear forces at 11/12/deck are compared to horizontal shear resistance provided across that plane by concrete friction and rebar.

The relevant page appears to be: UCPP_Final_Calculations_Superstructure (1).pdf PDF Page 1284 (labeled 1283) where there are three tables listing and calculating several variables for each of the deck-to-member nodes.

It's a bit of a chore to understand all the variables, but they are described on pages PDF 1295 and following, though the example there is for vertical shear planes, and page 1284 is about horizontal shear planes.

(I also found a PDF 2012 copy of the AASHTO bridge manual which covers the same material, and is searchable for variable names:
http://www.ddzph.com/wp-content/uploads/2017/03/AA...)

Description of variables and reasoning:

bvi: width of interface (21")

Lvi: Length of interface. For most of the nodes this matches the lengths shown on the drawings on PDF pages 1440 and 1441. Oddly, the lengths used for the end nodes are conspicuously shorter than the length on the drawings, but do correspond to the length over which the rebar is distributed, plus a modest amount. That oddity would underestimate the cohesion component of stress resistance, but that's zeroed out by c = 0.

Acv = bvi x Lvi Area of interface

Pc: Compression force perpendicular to the shear interface. So in this case, the vertical component of Compression force in the diagonals. For the end diagonals 2 and 11, we would expect this to be half the bridge weight, or 950 kips. For 2 and 11, Pc is overstated at around 1250 kips, which I will come back to.

VDC, VLL, VPT, VTU+TD. These are apparently components of horizontal stress. VDC (stress due to dead load of structural components and any nonstructural attachments) looks extremely understated for members 2 and 11. Given a vertical load of 950 kips, and member angles less than 45 degrees, the horizontal component VDC has to be more than 950, not 589. And if Pc is supposedly around 1250, then VDC has to be more than that. (And in previous calcs I have shown it to be even higher.)

Vui: This is a crucial variable -- the total horizontal stress to be resisted.
Vui = the sum of VDC, VLL, VPT, VTU+TD multiplied by their respective load factors (provided in the "Load Factors" table.)

Needless to say, if VDC is understated, then so is Vui, by an even larger amount since DC load factor is 1.25.

Avf: Area of rebars passing through the shear plane. Avf tells the number of rebars, which are all #7 whose cross section is 0.6 sq-in. These correspond to the bars seen on PDF page 1448. None of the rebar area Avf figures look correct, and several of them are negative!

Since one of the purposes is to determine whether the rebars are sufficient, the fact negative rebar areas were not noticed seems to me a rather significant indication of the level of scrutiny this vitally important calculation underwent.

Lest we disbelieve that Avf is maybe not what I just described, continue on to the third table on the page.

Here the crucial variable is Vni, Nominal Interface Shear Resistance. One of the most crucial design checks is that shear resistance is greater than shear force, or Vni > Vui.

Vni = c x Acv + mu (Avf x fy + Pc) where

c (cohesion coefficient) = 0 for this analysis (making Acv moot, as previously noted)

mu = coefficient of friction = 1

Avf = area of rebar passing through the interface

fy = 60 ksi (shear resistance of rebar per area)

Pc = force normal to interface = 1233 for member 11. So note that overstating Pc by 30% means the shear resistance will be overstated by 30%

So Vni = 1 (-2.28 X 60 + 1233) = 1096.

Note that this matches the third table, showing that the calculation blithely subtracted the shear resistance of the negative rebars, verifying that Avf is indeed what I surmised!

The negative rebars error understates the shear resistance, hence would have made the bridge safer, but it also shows that effectively the shear resisting rebars were simply not calculated at all. So the calculations supposedly "validate" that all the shear resistance can come from the concrete friction resulting from the normal force, with no rebars.

The following image shows several versions of the calculation of the shear force (Vui), and the implications when compared to Vni (shear resistance), and resulting rebar requirement.



Of note:

1. So far as I can tell, the bridge should have been designed according to the "factored all" row, in which case the provided 4.8 sq-in of rebar is 15.4 sq-in short of the requirement.

2. But as a better idea of whether the bridge should be able to stand up at all, we might neglect the live load, and the load factor, and just use raw VDC as the horizontal shear. That would result in a requirement for 9.7 sq-in of rebar, double what was provided. (The calculations page conservatively omits any reliance on concrete cohesion, so presumably there some additional shear resistance from that source not counted here.)

Summary:

• The horizontal shear force VDC was stated as 589, clearly less than the 950 it would be for a 45 degree member (ie: no-calc reality check should fail), and hugely less than the 1531 I calculated for #11's actual angle of 32 degrees.
• The horizontal shear resistance concrete friction component was overstated by 30% by overstating Pc. (But that error was reduced by the contribution of the "negative rebars".)
• The rebar area was negative, which essentially means that so far as the calculations are concerned, the rebars were unnecessary for shear resistance, a clearly absurd result.

These problems appear so conspicuous that it seems hard to believe that these are the calculations actually used, or that these issues went unnoticed. Hence my interest in having someone with a structural background take a look.

An undergraduate engineering student from my time as a student, without any computing device, would easily have been able to arrive at the following result on the back of an envelope:



I think this tells us what has happened to college education in this field since the introduction of computers. Nobody today can "see" where the maximum forces and stresses in a structure are located. Engineers today wait for the color coded printout, without understanding that a back of the envelope calc. MUST be done in all cases to check that the order of magnitude is correct.

Tragedies will continue until colleges go back to teaching the old method of visualization and hand calculation of structural forces and stresses.

The calculation below took less than 10 minutes. The calculation is an "order of magnitude" check. It ignores live load, to provide a rapid check for absolute minimum steel required to tie diagonal 11 load to the deck load.
The calculation confirms the calculation above almost exactly. One could add factors and the like, but the correct answer is somewhere close to 28 square inches of steel needed to tie the compressive forces in diagonal 11 to the tensile forces running in the deck. NOT the 5 square inches actually provided.

Quote (FortyYearsExperience)

... 23.4 sq-in...

Thanks for commenting. As it happens I was in the process of tidying the spreadsheet I used, and discovered that I had applied the load factor for VDC twice. I have corrected that error in the post you commented on.

The general conclusion is the same. Just that the 23.4 sq-in should be "only" 15.4 sq-in.

And I think your sketch assumes that shear resistance is provided only by the steel, at 53 kips/sq-in, whereas the calculations page (and my calcs) get 950 kips of resistance from concrete friction under load of 950 kips, and use 60 kips/sq-in for the steel.

For what it's worth, 950 kips would be equivalent to 16 sq-in of steel. So that puts us both in the same ballpark.

Thanks for your interesting and thoughtful observations.

I believe however that the shear force between the tie and the deck is 1,450 kips, not 950 kips. (= one half weight of bridge, times cotan 32 degrees).

Also, I am puzzled that you find a reduction in shear can permissibly be attributed to concrete friction. The way I understand the shear capacity of reinforced concrete is to ignore the concrete when analyzing the steel. Then, do a separate check to see if the concrete can withstand the compression. In this case, I believe it cannot withstand the compression, but my check ends with the steel because it is hugely inadequate.

Appreciate any further thoughts.

Quote (FortyYearsExperience)

I believe however that the shear force between the tie and the deck is 1,450 kips, not 950 kips. (= one half weight of bridge, times cotan 32 degrees).

First, I think the 950 kips you're looking at is for Vni (resistance) not for shear force.

So the figure for plain horizontal force that I used was 1531 (not 1450), as follows:

I actually did it in two steps: 950 kips vertical --> 1802 kips diagonal --> 1531 kips horizontal. (This way I could match the diagonal force with a slide in the FIGGs "crack meeting" presentation).

However, to go directly:

Angle = 31.82 deg

Cotan(31.82 degrees) = 1.6115

Vertical force = 950

950 x 1.6115 = 1531 horizontal shear force.

Quote (FortyYearsExperience)

Also, I am puzzled that you find a reduction in shear can permissibly be attributed to concrete friction. The way I understand the shear capacity of reinforced concrete is to ignore the concrete when analyzing the steel.

I have very little background in this kind of analysis for actual materials. I'm just following what I recall from a first course in static forces, plus the calculations laid out on the referenced calculations page 1284, plus the description of the method in the AASHTO manual excerpt on pages 1295 and following.

So all I've really demonstrated is that the intended calculations suffer from wrong input data and wrong execution of the formulas that should have been calculated, producing results that seem so wrong (like negative rebar area) that they should have immediately triggered alarm.

I am not qualified to have an opinion on whether this was a suitable approach in the first place.

I think I have learned in this exercise that the shear resistance is seen to be composed of three parts:

1. Rebar shear resistance.
2. Friction at the shear interface (perpendicular force x friction coefficient)
3. Cohesion of the concrete, which is apparently not well characterized and thus controversial (and in calculations under discussion here, omitted).

I could certainly see how steel is the only component useful in tension. And maybe for assessing shear, there's a very conservative approach that again assumes the entire shear resistance would come from steel. But apparently that's not the approach taken here.

FIGG's calculations are extremely opaque, and this is evidently one of the sources of the problem. They could not even explain to themselves, in a checkable form, what they were calculating and why.

Quote (FortyYearsExperience)

FIGG's calculations are extremely opaque, and this is evidently one of the sources of the problem. They could not even explain to themselves, in a checkable form, what they were calculating and why.

Well I found the calcs a bit opaque at first. But after looking up what all the variables mean, it seems straightforward. It's also the same set of steps used on pdf page 1299, and also on the slides in the FIGG "cracks" meeting, albeit examining different shear planes. I infer that people who work with these variables everyday would find this easy to follow.

Though it is curious that the cracks meeting presented an examination (meeting PDF slides page 27 and following) of a different shear interface, rather than using the ones for which calculations were already available. (And as I've discussed earlier, the slides appear to overstate the shear interface areas, the perpendicular ("clamping") force and the rebars crossing the interface.)

I would even question the use of the 4 - #7 stirrups as contributing to the area of shear steel, the way they are detailed. The requirement states that the steel should cross the expected shear plane perpendicularly, AND that the steel be fully developed on either side of that plane. It's not clear from the design drawings how far the stirrups extend above the top of deck. Simply extending them above the deck construction joint (1" or 2" or 3") is not enough. To be fully developed (90-deg. bend) they'd need to extend at least 11" above the deck. And, if you wanted to use the steel coming from the diagonal for this purpose (that's assuming those J-bars are placed properly and fully developed), because they're crossing the shear plane at an angle, I would reduce the effective area based on the angle of the incoming diagonal--say use only 32/90 of those bars.

Is it clear where FIGG is pulling the numbers from for doing these calcs? I've looked through parts of the calcs only briefly. Page after page of tabular data cause my eyes to glaze over. Any thoughts on why 2 computer models were done. I haven't looked at the FEA at all, except to note that it's undated, done by someone else (not the LARSA modeler), and models only stage 1 of construction.

Quote (TheGreenLama )

I would even question the use of the 4 - #7 stirrups as contributing to the area of shear steel, the way they are detailed. The requirement states that the steel should cross the expected shear plane perpendicularly, AND that the steel be fully developed on either side of that plane.

I entirely agree. There is, in reality, zero amount of steel provided to tie diagonal 11 to the deck.

Reinforced concrete beam resisting bending - BE of Reinforced Concrete with Prof Ibell Pt2

https://youtu.be/GPOAelSOLFs?t=69

Insightful explanation of how to control failure mode of reinforced concrete beams.

Has anyone here seen any evidence the emergency center-line shims were actuality installed and when? Is it possible the shim jack was being operated in conjunction with the PT rod operation? Obviously the diagram would have to be lifted some order of magnitude to fit the shim effectively.

Bear with me for a moment... I have faith in Figg in so far as engineering goes. I argue they think just like us and it rarely is useful to underestimate people. When they decide it's necessary to shim center-line, implement an "exoskeleton" to add confinement to 11/12 node, and get CIP tied in ASAP, I believe them on that. These recommendations suggest they knew what was wrong even if candor may have been lacking to some extent.

It's my opinion most structure engineers and construction managers would have resisted group-think and taken action to get road closed immediately upon examining that diagonal no matter what. It's just by fate none on this project quite reached that level. Possibly because closing the road was paramount to realizing structure was doomed and un-repairable.

GWIDEMAN: Here are some excellent lectures to assist in your quest.

https://www.youtube.com/channel/UCIL0NyWPPAqrikLb6...

Video lectures are by David Garber, PhD, assistant professor at FIU (coincidence, link not submitted to be provocative)

I must have missed this earlier, but FIU is a research hub for ABC - Accelerated Bridge Construction - and does research sponsored by the US Dept of Transportation. David Garber, mentioned by jrs_87 above, is an Assistant Professor of Structural Engineering and a member of the Accelerated Bridge Construction University Transportation Center.

Whenever "FIU" has been mentioned in conjunction with the design of this bridge, I just assumed it meant the university administration, but now it looks like the bridge may have been intended to be some sort of recruiting showpiece for the FIU Engineering Department. In any case, I wonder how many of their faculty have left after this debacle.

EDIT ADD: This Miami Herald Article from 2018 March 15 reported that a university spokesman said, just after the span was installed, that the FIU ABC center was not formally involved in the bridge project. Below is a section from the article:

At this point I'm fairly satisfied that, as described in my earlier post, UCPP_Final_Calculations_Superstructure (1).pdf PDF Page 1284 (labeled 1283) is where the stress resistance is supposedly checked against the stress demand, and wrongly passed due to wrong input numbers.

So I agree with:

Quote (TheGreenLama)

Is it clear where FIGG is pulling the numbers from for doing these calcs?

... that one point of interest now is where did those numbers come from?

In summary, the numbers that I think are of most interest are the red ones in the following table:




Here I show what I think are the correct values for key variables, and the four calculations that we've seen presented. (I've included the calculations pertaining to member #2, because it allows comparison to the handwritten notes page.)

Clearly the biggest puzzle is how the half-weight of the bridge was not the number used in the key calculations. The angles of the members were also wrong with a comparably harmful effect on the calculations.

A related issue is how the rebar area on page 1284 came out negative, which is impossible. That didn't help the bridge to pass muster, but it does imply incorrect formulas behind the result table, and thus a flawed process in setting up this critical design-check report and also the checking of it, both of which would make illuminating lessons.

Some general thoughts on approach and speculation on load disparities.
Two different shear planes, two different approaches to design.

Shear Plane 2 (as imagined on p.1298 of calcs)

External loads are applied to hand-drawn free body diagram.
Easy to visualize. Easy to get a feel for relative load magnitudes.
Presume loads are taken from 2D LARSA.


Shear Plane 1 (as shown on p.1381 of calcs)

Member 11-12 area is modeled as a 3D block in the FEM.
From within this concrete block a "slice" is defined.
Loads are extracted for this "slice", taken as gospel, and used for design.
No attempt made to gauge relative magnitudes of extracted values to real world numbers.

From the table on p.1283, Acv (concrete shear area) = 882 sq in, with Lvi = 42 in
This 42 in corresponds only to the area where the #7 stirrups are situated.
Could it be that Shear Plane 1 was checked using only the load over this area, excluding the Member 12 area?
Could it be that Shear Plane 1 was checked for fixed pylon case?
Could it be that the unaccounted for Member 11 load, which needs to be designed for, is just not crossing this plane slice due to modeling irregularities?
[Edit: Using this internal slice method to extract loads, is it conceivable that the longitudinal PT in the deck is impacting (i.e. incorrectly reducing) the design shear load on this internal slice?]

And, Gwideman, I agree that when looking at this puzzle it's good to incorporate Member 2 in the summaries, for comparison purposes.

Quote (TheGreenLama)

p1298 of calcs... Shear Plane 1... Presume loads are taken from 2D LARSA.

(aka PDF page 1299)

I agree that the free body diagram is satisfyingly visualizable. But it uses a vertical load of only 674 kips (and correspondingly low axial force).

Quote (TheGreenLama)

p1381 of calcs ... Could it be that Shear Plane 1 was checked using only the load over this area, excluding the Member 12 area? Could it be that Shear Plane 1 was checked for fixed pylon case?

(aka PDF page 1382)

I wondered those points too. But the vertical load on #12, while substantial, would surely be only a small fraction of the 950 kips half-weight of bridge.
And somewhat countering the pylon speculation is that the interface length chosen is only 5'-5.8", which includes the entire dashed red line, excluding the fixed pylon.

I did search the entire calculations PDF for the key numbers I've highlighted previously in red, and found none of them. (And when looking for "647" for example, I searched for "647", and also "646.", on the basis that the page 1382 calc might round up.)

Member 11 is offset from diaphragm support. Would areas I marked in orange also be subject to shear? Does anyone know where the steel for this is detailed? And would not the PT bars have messed with this?

Also, does not the offset of 11 and diaphragm greatly reduce vertical vector of clamping force against shear in yellow area? The diaphragm on south size was much larger because it did not share space with CIP wire stay pylon/tower.

Quote (jrs_87)

Member 11 is offset from diaphragm support. Would areas I marked in orange also be subject to shear? Does anyone know where the steel for this is detailed?

I guess the thinking is that almost any area is subject to shear. The "shear interfaces" looked at in analyses are not necessarily actual physical joints (but could be), but rather just places where the engineer thinks there may be a high level of shear stress, and thus a place to make some calculations. The chosen interface is usually (always?) flat, to make the calculations easy. I would certainly agree that the vertical line you drew would be a good place to analyze. The only drawings I've seen that detail that area are B-60 and B-61, and maybe B24A.

Quote (jrs_87)

And would not the PT bars have messed with this?

Which PT bars were you thinking? Axial in member 11? Not sure what to think about those.
Longitudinal in the deck? Those are off to the sides of the 11-12-dec connection.
Or maybe the transverse PT bars across the deck? In the drawing you used (from B-61), there are two oval-ish symbols just below the red/yellow line. I believe those are the channels for the transverse PT bars. But this is the only drawing where there's such a transverse PT bar located where your vertical orange line is. Other drawings, for example B-60, omit this endmost PT bar.

gwideman, I agree we should think that shear can express itself almost everywhere. I was just looking at load path of node and diaphragm and it vaguely reminded me of Hyatt hanging rod. But I digress until photos or drawings are released that show exactly where deck and node separated.

I was thinking about temp PT bars in 11. I knew transverse strands pass through node, thanks for telling me about location (ovals). Thanks for your post, I will study it more now...

PS. After watching some lectures I can see how negative values for re bar area are legitimate (assuming the re-bar is placed anyway).

Quote (jrs_87)

PS. After watching some lectures I can see how negative values for re bar area are legitimate (assuming the re-bar is placed anyway).

Really? That's intriguing. Which lecture(s) pertain? BTW, I did look at some of the links you posted previously -- thanks for those.

Quote (jrs_87)

I knew transverse strands pass through node, thanks for telling me about location (ovals).

... but according to other drawings, that last transverse PT strand is omitted and does not go through that 11-12-deck node. (Which I noted in a previous post would mean that the FIGG "crack meeting" presentation overstates the clamping force on #11's extension through to the end diaphragm.)

Quote (jrs_87)

Member 11 is offset from diaphragm support. Would areas I marked in orange also be subject to shear? Does anyone know where the steel for this is detailed?

It occurs to me that your vertical shear interface might be seen as "safe" immediately as follows:

They are using a coef of friction mu = 1. Since the angle of #11 is less than 45 degrees, the horizontal force, (and thus the friction), must be greater than the vertical force.

And the flipside of this same reasoning is that it would be obvious that for the horizontal shear interface we've been discussing, a significant amount of rebar would be required.

Quote (gwideman)

Really? That's intriguing. Which lecture(s) pertain?

(re: negative rebar area)

Sorry, I don't know yet exactly what I saw and where. Perhaps we can get more specific and compare notes later. The NTSB report is months away so plenty of time remains. The idea was iterative adjustments and checks with the primary goal of not making structure stand, but determining how it would fail i.e. provide warning. Figg seems to have used same codes in presentation to present a case structure would not collapse because it passed design checks. But the codes don't mean that exactly. Figg had the warning the code provides but seems to have discounted it.

You may have seen in lectures where, for the sake of simplified equations, steel (in the equation) is replaced with equivalent (more area) concrete. Side point> We have all heard concrete is strong in compression and weak in tension. One Phd said that is wrong, it is weak in both. I'm losing focus and need to take a break for a week or so.

Quote (gwideman)

They are using a coef of friction mu = 1. Since the angle of #11 is less than 45 degrees, the horizontal force, (and thus the friction), must be greater than the vertical force

I have to make a point here. Nowhere, absolutely nowhere, in reinforced concrete design does "friction" come into any analysis. Concrete does not provide any frictional resistance. It is not a "thing." You will not find friction mentioned anywhere in any learned treatise on reinforced concrete design. The only structure that can resist shear or tension in a reinforced concrete structure is steel. In this case, no steel was provided to restrain the diagonal #11 from moving to the north (right, in the figure above), away from the tensile force that was trying to develop in the deck cables to move to the left. Those stirrups you see are not long enough to develop any tension load within the concrete, and even if they were long enough, they are totally inadequate in sectional area. This has been demonstrated above.

I'm assuming you are discounting the Shear Friction method described in ACI 318 as "not really friction"?

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Quote (FortyYearsExperience)

I have to make a point here. Nowhere, absolutely nowhere, in reinforced concrete design does "friction" come into any analysis.

AASHTO LFRD Bridge Design Specs 2014 page 5-82 and following (included as PDF pages 1297 and following of the FIU "Final calculations" PDF):

Quote (JAE)

I'm assuming you are discounting the Shear Friction method described in ACI 318 as "not really friction"?

JAE, maybe FortyYearsExpereince is hokie66

Quote (gwideman)

the interface length chosen is only 5'-5.8", which includes the entire dashed red line,

Just to finish my thought. This 5.49' is shown on pp.1387 & 1381 of calcs (as 9.598 sq ft or 1382 sq in). It includes both members 11 & 12.
On p.1283 of calcs, the Acv shown in table is only 882 sq in. No "LUSAS View" slice drawing is shown for this. I'm assuming they're referring to the same plane, but here instead of 5.49' they only consider 3.5'.
Thus, missing area would mean missing load. At least that was the thought.

Quote (gwideman)

AASHTO LFRD Bridge Design Specs 2014 page 5-82 and following (included as PDF pages 1297 and following of the FIU "Final calculations" PDF):

Precisely. This code tells you how to calculate shear resistance via frictional force transfer between surfaces of steel or concrete that are cast or placed as separate or joined entities AND THEN CLAMPED TOGETHER. By contrast, my point relates to poured concrete reinforced with internal steel rods or cables, like in the bridge that collapsed. Nowhere in the design codes for such reinforced concrete does internal concrete "friction" provide any credit for reducing the amount of steel that must be provided to withstand tensile shear forces within the structure. As you can see, THERE ARE NO CLAIMPING FORCES in this bridge structure (and certainly not in the vicinity of the failure cracks that were observed). So, I am not "discounting the Shear Friction method described in ACI 318 as 'not really friction'." What I am asserting is that in rebar reinforced concrete design (where there are no clamping forces), concrete friction is not a factor for reducing the steel required to withstand tensile shear forces. If in fact the design of the FIU bridge relies on "friction," internal to the concrete poured to make up the bridge, in order to reduce the shear steel requirement, then that would certainly go a long way to explain why it collapsed.
See for example the following table which exemplifies the kinds of interface where it is appropriate to consider friction between opposing surfaces. Category 4 is not rebar reinforced concrete.

Quote (FortyYearsExperience )

See for example the following table which exemplifies the kinds of interface.

Sorry, I'm not clear on the point you are making. Are you saying that this table covers the case of the FIU bridge? (If so, which of the listed Cases apply?)

Or that the table covers cases where friction applies, but that it does not cover the FIU bridge? (If so, why does it not apply?)

Thanks.

Quote (gwideman)

I'm not clear on the point you are making.

The table covers cases where friction applies, but it does not cover the relevant part of the FIU bridge because the shear plane that had the highest shear stress in the FIU bridge (namely the condition that is the subject of my 'back of the envelope' calculation above) did not include any clamping force. The mere suggestion that any code permits reliance on a "friction" force supplied by concrete to reduce the tensile shear steel demand in such case is nonsensical.

Quote (FortyYearsExperience)

The table covers [...] does not cover the relevant part of the FIU bridge because the shear plane [under discussion here] did not include any clamping force.

As I understand it, the "clamping force" scenario is where the interface is rough to some amplitude, so that if shear is to occur, the mating bumps and valleys would have to ride up on each other, which would create a separation across the interface and thus meet the tension force of the rebars that cross normal to the interface. So the rebars would be designed to supply adequate tension (as oppose to sheer per se), and also must be sufficiently developed. It is that tension that's referred to as "clamping force"

On brief googling, whether or not to add externally applied force as part of the force normal to the interface (resisting the "riding up") is a subject that appears to have developed over the years. For example:

"Examination of shear friction design provisions" http://scholarsmine.mst.edu/cgi/viewcontent.cgi?article=8759&context=masters_theses
... notes:

"The equation for the nominal strength of the interface given in the AASHTO LRFD [2016] provisions is provided in Eq. 2.5. [...] The design for interface shear transfer in the AASHTO LRFD provisions is quite different compared to the ACI 318 code and PCI Design Handbook [...]. Pc accounts for the addition of any normal force that is applied to the interface (compression is taken as positive in the equation). Pc is added to the clamping force, Avf x fy, since both forces are acting normal to the shear interface plane. [...]
Vn = c Acv+ mu ( Avf fy + Pc) (Equation 2.5) "

I'm not versed enough in this discipline to know whether this is good, bad or indifferent. But it does look like it's an accepted idea in some circles, and corresponds to FIGG's calculations, albeit with (I believe) the wrong input numbers.

Quote (gwideman)

This academic thesis makes it quite clear that the notional "friction" produced in the concrete is a factor that allows for greater shear stresses to be accommodated in the concrete. This analysis is silent on the question of any reduction being allowed in the required tensile/shear steel. As my "back of the envelope" calculations show, the tensile/shear steel provided in the bridge to tie diagonal #11 to the deck was wholly inadequate. About 27 square inches of tie steel was required (absolute minimum). Whereas, 5 square inches of stirrups were provided, and they were not long enough to develop any tension force. The FIU/Figg calculations referred to "friction" in the concrete, but that is irrelevant to the question of required shear steel.

Quote (FortyYearsExperience)

minimum steel to tie A to B = 1.45E3 [kips]/53 [kips/sqin] = 27 sqin

Just to clarify for me: This is 27 sqin cross section of the rebar that should be placed perpendicular to the shear interface, right? That being the horizontal shear plane depicted in my 13 May 19 22:48 post ([Part 2] Now for the 11/12/deck analysis from UCPP_Final_Calculations_Superstructure (1).pdf page 1388.), right? If so, I'd like to understand this formula.

The shear interface has to connect the horizontal northward component of the member 11 axial force ultimately to the southward tension of the deck's tension rods. But being perpendicular to those two main forces, that steel is not directly in tension, but rather it would be in tension as a result of the interface trying to separate by the two faces "riding up" over the interlocking bumps and valleys.

So the calculation that involves the main force (your 1.45E3 kips) and the tensile strength of rebar (your 53 kips/sq in) would involve a factor to account for the riding-up mechanism. So I thought (but await your confirmation) that you are using:

Vn = X ( Avf * fy) or solving for minimum steel area: Avf = (1/X) * (Required Vn) / fy

Where:

• Avf: Area of rebar [sq in], to be found
• Required Vn: 1.45E3 [kips]
• fy: 53 [kips/sqin]
• X = either mu, or mu * lambda, where various tables show mu and lambda both having values of about 1.

In short, I'm trying to understand whether you are effectively using the same formula as FIGG, but using only the Avf * fy the term, and assuming X=1, or whether you are using something completely different, or even looking at a different shear plane.

gwideman: Shear and rebar. There may be more than one type of shear. One type is diagonal shear in a beam (the term diagonal means in the beam, not that the beam is necessary a diagonal member). An rc beam does not always require steel for shear, but it almost always needs longitudinal bar below neutral axis because concrete opens in tension too soon before it crushes in compression. Diagonal shear reinforcement (stirrups) are aligned at 45 degrees from stress or strain (not sure which) to resist shear. http://assakkaf.com/courses/ence355/lectures/part1...

Now, the type of shear you are looking at for 11/12 is related to resisting beam kick-out. I have no idea how that is quantified, but if the steel is perpendicular that would just be a "dowel effect" for which concrete crushes easily around the thin bars. Rebar is designed to resist tension along it's entire length, hence the rings around the bar. So I would expect some type of diaphragm to be required to resist kick-out. If friction is a factor, it becomes mute when crushing starts (little marbles of crushed concrete).

When I first looked at FIU bridge plans, I expected to see something like this:

https://www.youtube.com/watch?v=HjSvH9QLbkk

From our old youtuber friend in PA.

Audio recording allegedly made of Linda Figg pitching her concept to FIU. Difficult to listen to (excessive grandiose buzz words: award winning, iconic, signature, grand, one-of-a-kind...), but provides hindsight into what went wrong. (I'm not placing any blame on her, that would not be honest and fair at this point, there is more to this story.)

Quote (gwideman and jrs_87)

"There may be more than one type of shear."

Guys, time to get real. Can anyone commenting on this problem explain to me how they applied the concept of Mohr's Circle in resolving the forces that existed on the day of the failure? If you can't, then, really, you will not understand why the bridge collapsed. Evidently, FIGG did not consider this either.

In reality, a "shear" force is only a combination of tensile force and compressive force on a body, in which the angle of orientation of the two principal forces (tensile and compressive) may change at different parts of the body. There is no third force called a "shear" force once all the tension and compression forces are accounted for. Mohr's circle provides a perfect explanation of this principle. In reinforced concrete design, the tensile forces are taken by the reinforcing rods, and the compressive forces are taken by the concrete and the reinforcing rods. In the case of the failure of this bridge, there was effectively zero reinforcing rods to withstand the tensile component of the "shear" force existing between element #11 and the deck. End of story.

Hear,hear!
I spoke with Occam's duck and he assured me the guy was a quack, but you can't deny the obvious!

https://www.youtube.com/watch?v=HjSvH9QLbkk

This youtube posting is most revealing. Go to minute 37:23. Linda Figg pronounces that "It's all post tensioned concrete, . . . so everything is in compression."



As the poet said, "A little learning is a dangerous thing. . . " And yet, because we know that Linda Figg is not an engineer, she must have picked this idea up from an engineer on the team.

Common practice for design of post-tensioned concrete box girder bridges is to size concrete cross-section, PT, and cable drape such that concrete section always remains in compression due to axial and bending stresses.

Quote (FortyYearsExpereince)

And yet, because we know that Linda Figg is not an engineer, she must have picked this idea up from an engineer on the team.

I am not sure if you are being facetious. If not, according to her Linked In profile she graduated from Auburn University in 1981 with a BS Civil Engineering, Structural.

37yearsexperience

If anyone is interested, you can view history of hokie66 and FortyYearsExperience (view history by clicking their handle) debating force terms: tensile, compression, shear, and moments. I do not know why it is such a touchy subject. Call me strange, but I enjoy the banter. Actually, there are no such forces, only 42 from which everything else is derived. So when forty years becomes 42, all will be enlighted. https://en.wikipedia.org/wiki/42_(number)

Quote (TheGreenLama)

Common practice for design of post-tensioned concrete box girder bridges . . . such that concrete section always remains in compression due to axial and bending stresses.

Funny how huge tensile cracks showed up in the FIU bridge's concrete deck directly after being loaded up under gravity. Also, funny how the FIU bridge is not a box girder bridge.

Quote (Ingenuity)

. . . according to her Linked In profile she graduated from Auburn University in 1981 with a BS Civil Engineering, Structural

Funny how Linda Figg cannot place a licensed Professional Engineer ("PE") qualification behind her name. This is a professional certification/licensing requirement in every state for a person selling professional engineering services to the public. Funny how the investigation has not touched on this point.

Quote (FortyYearsExperience)

Funny how Linda Figg cannot place a licensed Professional Engineer ("PE") qualification behind her name.

As the owner of a multi-billion dollar engineering firm, she's obviously far too busy to design anything, so she likely let her license expire. She wasn't the designer for the FIU bridge, so what licenses she holds or held is irrelevant. Her firm has plenty of licensed engineers to do the actual engineering work, which is obviously all that is required in the states where Figg has designed bridges (which is the majority).

Edit: "They [Figg] have built and managed bridges in 42 states and six countries..." Link If there was a requirement that the owner of the engineering firm be licensed to practice engineering in those 42 states, then you can be sure she'd be licensed in those states.

Quote (HotRod10)

so she likely let her license expire.

Funny how a search of the Florida database on Professional Engineers in the historical record of the state shows that she was never licensed. Although, the record does show her father was licensed, deceased in 2003. See below.

Also, funny that, without a professional license, she is heard selling to the public why the bridge is safe because "everything is in compression" -- when clearly that was false information. The point I made, above, is that somebody on the design team with a PE license clearly believed the same thing and fed her that false information.

Now, who on the record from the design team believed that the bridge was safe against all tensile forces? Could it be that the last minute frenzied tensioning of the steel cables in diagonal #11 was motivated by the belief that "more pretension in the steel cables" would put "more compression into the failing concrete." ? Seems like the investigation has gone silent on simple questions like this.

Making a sales pitch doesn't require a PE. Obviously, owning an engineering firm doesn't either; neither does commenting in general terms about a design, I believe.

Quote (FortyYearsExperience)

...a search of the Florida database on Professional Engineers in the historical record of the state shows that she was never licensed.

Did you check the registration lists in the other 49 states and the dozens of other countries that license engineers? Just because she wasn't licensed in Florida, doesn't mean she wasn't licensed. Anyway, it's still irrelevant, as there's no requirement that the owner of an engineering firm be a licensed PE.

I recommend this book: https://play.google.com/store/books/details?id=BAz...

Bridges: The science and art of the world's most inspiring structures
David Blockley
February 25, 2010
OUP Oxford

It is written by an engineer, is non-technical but does a good job acknowledging human challenges in bridge building. The chapters on history are very well done. This book is applicable to this thread, however, his writing style for this audience is sure to draw flaming attacks.

BTW, the audio linked and quoted above was recorded in 2015 - well before the design was done. So, the assertion that "she is heard selling to the public why the bridge is safe because "everything is in compression" -- when clearly that was false information." and the speculation that "she must have picked this idea up from an engineer on the team." are also a false assumptions.

Whie I agree with HotRod that her license status is not really relevant to her owning a compnay, it does sound really off for her to say "it's all in compression" when it clearly is and was not. She's representing an engineering firm that specializes in high-level bridge design, so she really needs to know better what she's talking about.

Back to the collapse though. I did a Google search to see if a beam with an open truss-frame web was ever looked at before, and I came across some research done on a beam of this type to be used for a roof. I haven't delved into too deeply yet, but it seems they had a lot of trouble dealing with the force concentrations where the struts met the flanges. They're section was also symmetrical about the midline, and they used consistent strut angles through out. They also had full prismatic sections for the 1st and last quarter (roughly) of the beam. The most telling (and obvious) statement of the paper was:

"The fact that concrete does not work efficiently in resisting tensile stress makes it very
difficult to design a truss made of solely reinforced concrete."

https://digitalcommons.unl.edu/cgi/viewcontent.cgi...

Here is a concrete truss type that indicates designer was concerned about tying members to deck.

jrs_87,

In that picture, tying to the deck is important, but that's not what's driving the increase in section as the post approaches the deck. Underneath the deck, at the posts, you can see a beam underneath. This is a "half-through pony truss", otherwise known as a "handrail truss" in my neck of the woods. While the posts and diagonals obviously brace the top chord at the panel points for buckling in the vertical direction, the posts rely on a moment connection with the beams under the deck to provide the out of plane bracing for the top chord (if SlideRule is still reading this thread, I imagine that he will see the obvious need to use Engesser's solution). So, any aggressive tying in to the deck is really a by-product of the posts tying in to the beams underneath.

Really, the deck is spanning between the beams underneath. There's really not going to be the crazy force transfer between the posts/diagonals and the deck as there was in Miami.

At any rate, that is a really cool looking bridge. Were you able to find a location for it?

Obviously as it ended up, in the FIU bridge, the concrete was not all in compression, which contributed to its failure. As designed, the concrete in post-tensioned and prestressed structures is usually supposed to stay in compression. On the occasions where tension occurs in the concrete, it is incidental and not included in the load resistance of the member.

I'm travelling for a few months, so haven't been monitoring this discussion. But just wanted to respond to Ingenuity's facetious comment about me perhaps masquerading as someone else. Not true, but funny. As to the concept of "shear friction", I still don't believe it is well thought out, but have said that many times...

It looks like a lot of people here still can't see the forest for the trees. A bridge built with concrete, even in the form we associate with a truss, inevitably acts as a frame, with all the attendant moments involved.

winelandv Re "Pony Bridge"

https://historicbridges.org/bridges/browser/?bridg...

Here it is, in California no less. It is actually a steel truss later encased in concrete.

Posters, question, EOR stated "strut" cracks would be repaired (with epoxy?). Is this yet another questionable assertion from Figg? What good would epoxy do in such a highly loaded member? And the second those deep cracks formed, did Linda Figg's promise of 150 year life span become undeliverable even if the bridge did not fall?

Brand new high-tech bridge collapse in Canada for which cause will never be found.
https://pressfrom.info/ca/news/canada/-93667-bridg...

"the bridge was built to Canadian standards"

"The Reeve of the rural municipality (RM) of Clayton says the bridge that collapsed six hours after it opened was built without having geotechnical investigation done on the riverbed it stood on. A bridge building expert calls that approach "irregular." "

https://www.cbc.ca/news/canada/saskatchewan/bridge...

Considering all the elements of the bridge remained unfractured, I'd say it sank.

"You can't drill through water," he said. "You can't do it. You can't take underground samples." (politician Hicks)

James Eads would like a word slipped in on that. His bridge which still stands on the bedrock below the muddy bottom of the Mississippi. https://en.wikipedia.org/wiki/Eads_Bridge

Looks like they are chasing leads on other bridges done in a similar way - https://www.cbc.ca/news/canada/saskatchewan/sask-g...

3DDave, Re: Canada Bridge
Upon further research, I have found this collapse to have extreme political overtones, like the FIU bridge.

Local politicians wanted to use inexpensive screw (helical) piles. Defend decision post collapse claiming no proof more expensive structure would have not failed as well.

https://www.cbc.ca/news/canada/saskatchewan/munici...

Helical Piles http://helicalpilebook.com/Files/Ch_01.pdf

(I'm probably starting to diverge off topic too much.)

Neither bridge failed for political reasons - they failed for fundamental engineering mistakes. I expect it is rare for any project to be touted as "spending extra money for no reason" so any claims about costs are universal and indicate zilch. The more common complaint is cost overruns. Likewise since most bridges are publicly funded for public use there are almost always political overtones. Unless one fails in a spectacular fashion neither the money nor the politics will get any note at all. Most people are so relieved that the construction is done they don't care.

The Saskatchewan bridge as been discussed here.
See:
thread815-444241: Rural Bridge Collapse Immediately After Opening

Bill
--------------------
"Why not the best?"
Jimmy Carter

For what it's worth, here's a member load summary extracted from the LARSA model. The issue with the calculations, however, is that they do not include any results for the construction stage at the time of collapse. As far as I can tell, the LARSA construction check done in the calcs only reflects the span during the move. As a result, the closest state I could find is the "end of construction" (EOC) conditions, which is NOT the load state on the span at the time of collapse.

So, a couple of things to keep in mind:
1) EOC state includes continuity PT, pylon, and faux stays.
2) It's presumed the PT case below excludes the permanent diagonal PT; see note at end.
3) Only two load cases are shown below. Others, such as LL, creep, etc., can be found in calcs.
4) A maximum load has been extracted for the LARSA member numbers shown below. These member numbers constitute the bulk of the length for each of the diagonals, top, and bottom chord members. As a result these loads are approximations only.
5) Loads shown are axial only (kips). Bending moments exist, but are not included in this summary. Axial loads are calculated by multiplying the "P/A @ centroid" by member area. If I'm interpreting LARSA output incorrectly please let me know.
6) These are uncombined loads w/o load factors.
7) Negative is compression.


LARSA Members:

Area (canopy) = 16.46 sq ft
Area (deck) = 45.67 sq ft
Area (diag 2) = 5.25 sq ft
Area (diag 3-11) = 3.5 sq ft


Self Weight Forces (kips):

Chord loads found on pp.579-583, diagonals on p.807 of calcs.
Diagonal 2 components (for 23.88 deg):
Horiz = 1653 k
Vert = 732 k
Diagonal 11 components (for 31.62 deg):
Horiz = 1155 k
Vert = 711 k
Total vertical = 711 + 732 = 1443 k

Rough hand calc of self weight of span based on input areas (excludes blisters and some incidental end zone area of each of the diagonals):
Clear span = 174.83' - 2.83' - 3' = 169' (this will exclude weight from Members 1 & 12 and end diaphragms, which are supported on the bearings)
Self weight of deck + canopy = (45.67 + 16.46) x 169' x 0.15 k/cu ft = 1575 k
Self weight of diagonals:
L2 = 29.1'
L3-11 = 162.8'
Wt = [(29.1 x 5.25) + (162.6 x 3.5)] x 0.15 k/cu ft = 108 k
Total self weight to be carried by truss = 1575 + 108 = 1683 k


Post-tensioning Forces (kips):

Chord loads found on pp.603-7, diagonals on p.811 of calcs.
Compare with PT stressing forces from design dwg B-69:
Total deck = (835 k x 10) + (527 x 2) = 9404 k
Total canopy (w/o cont.) = (531 + 534) x 2 = 2130 k
Total canopy (incl cont.) = 2130 + (556 x 4) = 4354 k
>>

Note that the two self weight loads, 1443 k for the LARSA model end diagonal forces, and 1683 k for the rough concrete volume, conflict with the 950 x 2 = 1900 k that has been quoted from the first post in this thread. Not sure if it's a result of my approximations, or if it's something else? The 950 tons was reported as the "lift" weight, which would include Members 1 & 12 and end diaphragms and additional 5.83' of span, so that may be where the difference lies.

Also, if you look at say Member 10, PT shows a positive value. This implies that diagonal PT is not included in the LARSA run for PT case shown above. Diagonal PT is sized on p.842 of calcs based on service load combinations, among others, for the critical axial and bending stresses. After having determined this diagonal PT, it's unclear if the model was then rerun to check long term stresses.

And finally, looking at Member 11 end loads. As Gwideman has pointed out earlier, the values used for calculating the shear reinforcement on p.1283 of calcs (Horiz = 589, Vert = 1233), which were supposedly pulled from the FEA, wildly contradict the loads found in their own complimentary LARSA analysis.

This is a first post for me. I have read from the first line to end. Not familiar with tools so have to just do it.
FYE 18 May 19 22:49
AASHTO and ACI allow the general formula used by FIGG to design reinforcing steel across an assumed and /or defined plane which transfers shear. This is called "shear Friction" - I did not coin the term. It is particularly useful in transferring shear from a deck slab to girders and designing corbels on columns. In the use of welded studs - "Nelson Studs" to the top flange of steel girders, the studs are inert and do not provide an active clamping force, but a 3/4" dia headed stud is allowed over 10 kips shear in developing composite action between the slab and steel girder. The concept of "shear friction" should work in the case of member 11 at the deck if correctly applied. I do have a concern about the amount of fractured aggregate I see in the photos, and whether a "mu" of 1 is perhaps too high for this concrete in this project. Also I have seen no confirmation of the preparation of the "cold joint" surface at the top of the deck.
In the use of Equation 5.85.1-3 of AASHTO 2016 (perhaps there is a newer one?) I do agree with FIGG that the "cohesion" should be disregarded, particularly in the case at hand.
" gwideman" has well described the intent of the influence of reinforcing across a "shear plane" (cold joint in this case) as used in the "shear friction" design concept.
Since the horizontal force at the base on member 11 is only present concurrently with a vertical component creating compression across the shear plane, it seems appropriate to allow use of the vertical component (unfactored) as a clamping force.
If the FIGG design complies with with the appropriate formulas for shear friction given in the applicable code (which allows reinforcing as across the joint a clamping force)and is properly applied to the particular joint , FIGG is off the hook for this item and codes need revision.
FYE "The way I understand the shear capacity of reinforced concrete is to ignore the concrete when analyzing the steel." Correct for designing a normally reinforced beam - when the shear stress exceeds a defined value, a crack is assumed and stirrups provided to support the shear at that section. Prestressed concrete design allows a contribution to shear resistance at sections which are under compression from the prestressing. The compression closes any cracking and maintains aggregate interlock. If one does not accept this concept, crossing a prestressed concrete bridge should be avoided.
The "10 minute calculation" (I use a lot of those - some not so quick) appears to develop the horizontal force of 725 tons in direct tension and tie that force back into the deck using longitudinal reinforcing of 27 sq in at 53 ksi. In this case the 725 ton force never made it across the cold joint.
I'm interested - what do you find the tension across the "cold joint" to be using Mohr's Circle? Member 11 is in compression, the deck is in compression from both longitudinal and transverse post tensioning. The existence of the cold joint defines the orientation of the tension developed by Mohr's - perhaps that analysis is more appropriate for a condition like this.
Regarding the development of the reinforcing across the cold joint as questioned by TheGreenLama, the dimensions for vertical placement should have been provided on the drawings. Perhaps they are on the rebar shop drawings.
Great forum - great questions - great responses - great format and tools.
Thank you.

Reinforcing across cold joints at deck surface
In reviewing the details of reinforcing which connects the diagonals to the deck slab, I see some interesting things.
First, a bit of background - in a simple Warren Truss, the diagonals transmit loads through a shear connection to the top and bottom flanges, creating compression in the top chord (canopy in this case) and tension in the bottom chord (deck). From the center (where shear is zero if the truss is uniformly loaded and symmetrical in geometry), each panel point (node) adds loads to the canopy and deck. The sum of these loads will add to become the horizontal shear at the bottom of the end diagonals in the case at hand. When viewing sheet B-61 titled DECK REINFORCEMENT & P T -MAIN SPAN (2 OF 2) and focusing on the #7 ties which resist the shear across the "cold Joint"s at the top of the deck, we see:
Truss Member 1 & 2 has 5 - ties "7S01"
Truss member 3 & 4 has 9 - ties "7S01"
Truss member 5 & 6 has 6 - ties "7S01"
Truss member 7 & 8 has 6 - ties "7S01"
Truss member 9 & 10 has 6 - ties "7S01"
Truss member 11 & 12 has 4 - ties "7S01" <<<<< the shear load is the sum of shears from nodes 7&8 plus 9&10 (plus node 10&11) but node 11&12 has only 4 ties instead of the 6 that nodes with less loads have. In effect the deck connection the north diagonal 11 is resisting loads generated by 12 ties plus node 10&11 and has only 4 ties to do so. Of course the vertical component of #11 reduces the steel requirement but ------- apparently not enough, judging by the results.
Without doing any calcs, it can be said "That's not right". Particularly at the end where breakout is a concern.
When watching the news on the west coast about 1:30 PM on March 15, I could see the bridge failed at the north end. Soon after it was learned that it was a concrete truss. Knowing that heel joints in trusses are critical, that became the suspect joint. Here we are.
I am saddened at the loss, and those lost and injured and their families have my sincere condolences.

Quote (samwise753)

Whie I agree with HotRod that her license status is not really relevant to her owning a compnay, it does sound really off for her to say "it's all in compression" when it clearly is and was not. She's representing an engineering firm that specializes in high-level bridge design, so she really needs to know better what she's talking about.

Back to the collapse though. I did a Google search to see if a beam with an open truss-frame web was ever looked at before, and I came across some research done on a beam of this type to be used for a roof. I haven't delved into too deeply yet, but it seems they had a lot of trouble dealing with the force concentrations where the struts met the flanges. They're section was also symmetrical about the midline, and they used consistent strut angles through out. They also had full prismatic sections for the 1st and last quarter (roughly) of the beam. The most telling (and obvious) statement of the paper was:

"The fact that concrete does not work efficiently in resisting tensile stress makes it very
difficult to design a truss made of solely reinforced concrete."

https://digitalcommons.unl.edu/cgi/viewcontent.cgi...

Great article. Though I would say that quote barely scratches the surface of the good and relevant material in it. Worth a read or even a skim look at the pictures for anybody following this thread. The examples shown show clearly that failure similar the Miami bridge are of key concern hence solid webs at the ends of the trusses. The ultimate failure of the test truss also bares similarities with the Miami bridge though at a much more reasonable load.

Upon re-reading HWY18MH009 NTSB report, I found a single mention of The Corradino Group, Inc. According to the report, this firm was involved in the FIU project. Does anyone here have any insight as to the probable nature of this firm's involvement?

I can't prove it, but they do transportation planning, and may have been involved in the bus stop planning...

SF Charlie
Eng-Tips.com Forum Policies

Re: The Corradino Group, Inc. - Did they perhaps do the traffic control planning for the move?

LOADS ON SHIMS, FIGG POWERPOINT PG 4 OF 44, 15 MARCH, 2018
The last paragraph states:
"The bending moments that develop in the continuous structure, when the falsework of the CIP Back Span is removed, will reduce the load on the shims from their current values."
I do not agree - in fact, I see exactly the opposite happening. Creating continuity over the interior pier and subsequently releasing the vertical load of the back span will leave a negative moment over the pier which did not exist at the time that statement was presented, and that negative moment will draw shear from the south span onto the intermediate pier, increasing the load in member 11 and adding load onto the bearing points.
Perhaps it was intended to grout fully under the north diaphragm when casting the filler concrete over the intermediate pier and that grout would support any increase in load at the pier? That does not sound like it will "reduce the load on the shims from their current values". It could share any subsequent loads which develop (after any shrinkage in the grout has been accommodated).

In a somewhat related issue, has anyone seen in the FIGG calculations an analysis of the vertical movements of the long span under changing temperatures or night/day differences in expansion or contraction of the upper flange (canopy) with respect to the bottom deck? It seems to me the canopy would cool much faster after sundown than the deck, causing a downward movement and subsequent strains in the web members and joints.

And one more, if I may. Was vortex shedding considered in the Figg calculations for the 16" pipe "suspension strands"? It seems to me to be a very real possibility. Imagine all those pipes vibrating at different frequencies. They will also expand and contract with changes in temperature and some are quite long.
Thank you.

Quote (jrs_87)

Upon re-reading HWY18MH009 NTSB report, I found a single mention of The Corradino Group, Inc. According to the report, this firm was involved in the FIU project. Does anyone here have any insight as to the probable nature of this firm's involvement?

According to the contract between FIU and Bolton Perez, The Corradino Group, Inc. was a subconsultant with the task of Structural Engineering/ Bridge Inspection. All the contracts (and other files of interest) can be found here: Link

It's worth repeating this link: http://facilities.fiu.edu/projects/BT-904-PRR.htm Especially 17. Time lapse videos (Caution File is large 16GB)

Digging around here you will find high-resolution time lapse videos that you may have seen before, but perhaps not at this resolution. I was able to see someone was indeed down at 11/12 spotting while PT work was being done at blister right before collapse. It would be very difficult to dispute tension was being added to get some type of mechanical result at 11/12. It is also easy to detect uneven lifting and rocking of span (not to say it was undue amount). And you can observe how many times in the two days before collapse, workers visited 11/12.

http://facilities.fiu.edu/projects/BT-904-PRR.htm
14. Recordings of Design Build team selection

This link has direct source of Linda Figg's pitch and other pitch from Facchina. Finally UCPP Deliberations 11.5.15.wav directly reveals FIU/Sweetwater efforts to get the fancy bridge they wanted at federal taxpayer expense.

I'd like to say that clearly, by re-stressing the temporary PT in Member 11, it was FIGG's intention to both 1) increase the vertical clamping force across the shear plane, and 2) reengage the PT rods in a wild attempt to quickly "add" steel across a failing joint that was suddenly found to be inadequately reinforced. But, sadly, I don't know FIGG's intention. No one does, yet. Prior to this reckless operation, the only things holding Member 11 to the deck were the Member 12/end diaphragm connection, and the unstressed PT rods laying loosely in the ducts of Member 11.

A question: If the PT rods were being engaged in shear at the time of re-stressing, would that have been apparent to the jacking crew?

Disclaimer: I just want to say that in no way do I approve of the above course of action.

Apparent actual superstructure designer of FIU bridge is one Eddy D Leon. He is fully qualified. https://www.linkedin.com/in/eddy-leon-p-e-b524b3a6

Quote from his calculations document:

"The nodal zones of the superstructure were designed for interface shear transfer between the interface of the diagonals with the deck and canopy. AASHTO LRFD 5.8.4 was used. Shear forces at the nodal regions were extracted from the LUSAS finite element model. The required size of the interface shear reinforcement was calculated for the critical region and provided conservatively to all regions." Highlight added by me.

The AASHTO LRFD Bridge Design Specifications: Section 5 Reorganization
by R. Kent Montgomery, FIGG, Dr. Shri Bhide, Bentley Systems Inc., and Gregg Freeby, Texas Department of Transportation

http://www.aspirebridge.com/magazine/2017Winter/AA...

GreenLama, I can't answer your question but I've often wondered what type of feedback the work crew was getting by means of their equipment. I wrote off thinking about it for lack of info on the procedure or monitoring of the structure. If the PT cables were pinched, it was at the far end so tension could still develop over the greater length of the cable and seem normal. I was thinking more along the line of the damaged #11 deflecting and not allowing tension to develop fully, i.e. just pulling the 10/11/canopy node down. Nothing would move much but the crew could be left scratching their head wondering why the tension wasn't coming up all the while pumping more energy into the structure. The reality, of course, was that this whole approach was just a pipe dream.

Some other thoughts:

From the Investigative Update, it is not unreasonable to suggest that #11 has shifted longitudinally w.r.t.



the deck or that #11 is no longer in contact with the deck for the purpose of engaging friction. My next question was what deformation is allowing these gaps? Is the deck sagging or rather how much? One thing is obvious, #11 is tortured and cannot be reliably modelled.

From the FIGG Presentation, it's acknowledged by FIGG that the photos do not even do justice for the gravity of the situation.





Is FIGG saying that they were up on the deck again just prior to walking into the meeting? This being almost two days after the photo above was taken? I really can't imagine what they were thinking. The presentation w.r.t. interim remedies is just bonkers, even the idea of reconsidering their analysis of an ideal structure in the face of obvious shortcomings defies rational thought when it comes to considering interim remedies. They just didn't want to lose face by admitting that a more obtrusive intervention was required and bullied everyone else with a hurry up offence while throwing caution to the wind.

But that's just my opinion.

Sym P. Ie, clearly the movement is longitudinal. From those other photos it's apparent that end tear out has already begun. What would be an interesting project for someone who's tech savvy in this area would be to generate a 3D model from those crack photos. Something similar to this: https://www.youtube.com/watch?v=BLxJGFzX0Qc. Maybe it's not possible with the few photos available, but if they could somehow be stitched together maybe we all could get that same "in person" sensation that FIGG noted in their presentation.

"No. 3 - Those involved in the deadly FIU bridge collapse case will head to court today. A judge is ordering the company who designed the bridge to testify over why they aren't able to provide investigators the text messages they sent or received about the bridge in the days leading up to the collapse." That's the whole item.

SF Charlie
Eng-Tips.com Forum Policies

One thing that has concerned me about this bridge from when I first heard about it: Holding the casting form diaphragm base to be exactly relatively coplanar to the permanent pier landings would be absolutely required, and yet difficult to achieve. What if the form supports settled during casting or were not built with this in mind to begin with? The entire surface area of span was supported during casting, so I admit settlement would not be expected, but it should still be kept in mind.

Underside of north side diaphragm shown here. Note what appear to be four through holes for the 190 109 foot pylon.

You've touched on one of the biggest unassuming issues with the placement and it screams from this perspective. The primary vertical load vector passes directly through the pad beneath #12 and yet there was no shim at this location. Instead the forces had to be carried through the slab, around those four holes to the laterally placed shims. This was apparently addressed early after the placement but I have not seen details of how this was carried out.

That was an oops they thought they got away with!

The lack of vertical support directly under node 11/12 was the main focus of the FIGG PowerPoint Presentation of March 15 meeting. They could not explain the cracking at the top of the slab so they proposed to place more shims directly under the node ( between the spaced bearing pads).
I think I recall a 1/2 degree limit for handling/moving the bridge. That is about 1-7/8" across the top of the pier. They would set forms much more level than that. Then there is the chance to level the tops of the bearing pads.
As to the outcome of the meeting, I am reminded of the Kenny Rogers song.
You got to know when to hold them
Know when to fold them
Know when to walk away
Know when to RUN!

For the record, $42 million to go victims of FIU bridge collapse in settlement.

- MCM is bankrupt
- FIU waives claim to $5 million professional liability policy held by MCM
- Families of victims end legal claims against FIU

Vance Wiley, thanks for your correction.

From FIGG's presentation, on page 1:

On 03/13/18, MCM e-mailed FIGG documentation regarding the cracks and FIGG
instructed MCM to install the recommended temporary shims in the pylon base directly
below member 12 (nodal area of members 11/12) between the permanent support
shims

from the slide on page 6:

Tuesday morning, upon seeing MCM'S information, FIGG requested that, as a prudent action MCM immediately install temporary shims directly below the nodal area of members 11/12 and the top of the Pylon/Pier, while further evaluations were on-going by FIGG.


Is there any indication that this was completed or were they still working up to it? Did they lift the span and reset it or just jam stuff in there? Did the remedial action relieve any relevant distress to the structure? Also, the presentation appears oblivious to the significant cut-outs. It just adds to the high drama of incompetence.

Quote (jrs_87)

Note what appear to be four through holes for the 190 foot pylon.

Quote (Sym P. le)

Instead the forces had to be carried through the slab, around those four holes to the laterally placed shims.

Regarding the four holes, and in connection with the Mar 15 "cracks" meeting and PowerPoint presentation. You might look back to May 11th on the current page of this thread, where this was discussed:



Basically the PowerPoint calculation of shear looked at the vertical planes on each side of the projection of #11 through the deck and diaphragm (not the horizontal plane between #11 and deck), and failed to take into account the four holes. So that area was overstated by 50% (ie: should be reduced by 33%). That was one of three aspects of the presentation's calculation that appears to overstate the shear handling capability of these planes.

Quote (TheGreenLama)

here's a member load summary extracted from the LARSA model

Thanks for posting those diagrams, which add significant understanding of the otherwise somewhat cryptic pages of tables in the calcs.

Do you have further insight into the meaning of the multiple columns (P/A@Centroid, And Normal Stress at points 1..4), and also the multiple rows per Member (Stations 0..2)?

Re: Sym P. le (Mechanical)5 Jun 19 21:20

Your mention of slides reminds me pdf states "first few slides were not photographed", so are missing from us.

As for the emergency shims, timelapse video covers that time frame. http://facilities.fiu.edu/projects/BT-904-PRR.htm 16GB

While not conclusive, I see no sign in video emergency shims were actually installed.

gwideman - I think you called it here - you posted "Basically the PowerPoint calculation of shear looked at the vertical planes on each side of the projection of #11 through the deck and diaphragm (not the horizontal plane between #11 and deck)".
From what I recall, a recent post (yours, I think) had a shear plane at the top of the deck until it intercepted the sloping plane of the PT rods and followed down at the PT slope, where it exited the end of the deck. I think that will be found to be the failure plane. It also explains the cracks in the top of the deck, which FIGG could not explain in pages 39,40,and 41 of their PP show. As the shear plane followed the PT slope it went deeper into the top of the deck, causing the cracks at the surface and rupture from the end of the deck.

VW - if the failure was in the bulk of the deck then that chunk of deck would have remained attached to 11/12. Instead the deck was sufficiently strong as to pull the lower PT bar out of 11. The failure plane is the cold joint between 11/12 and the top surface of the deck.

Gwideman, I'm no expert in LARSA, but this is how I interpret the output format. Each "beam" member input into the computer model goes between defined nodes. Station 0 would refer to the first node, station 2 refers to the end node, and station 1 is the midpoint.

For stresses you're following the basic formula (for 2D behavior):
normal stress = (P/A) +/- (Mc/I)
where P is axial force, A is area, M is moment, c is distance away from centroid to stress point, and I is the moment of inertia.

The "points" are then locations on the cross section where a stress is calculated. These points are defined in the input somewhere. For the rectangular diagonal members I believe they've been defined as the four corner points, beginning in the upper right corner, then going around clockwise. For the deck, where 6 points are defined (I believe 6 is the max allowable for LARSA), I'm not exactly sure. But they'll be extreme points on the cross section, and will move clockwise around the perimeter.

In our situation, where we only have bending in one direction, half the points are redundant. So, from this stress table you should be able to back out axial and moment forces.

Thanks TheGreenLama, very helpful.

3DDave - I agree the the failure likely began at the cold joint - and days before the collapse. But it appears that, on the way, something developed enough resistance to the movements to transfer enough loads to the top portion of the deck, and cause the cracking photographed on March 13, at the west side of the node and the beginning of the blowout in the end of the deck.
Thanks for your comment.

I uploaded an edited version of one of the FIU timelapse movies that is only concerned with the bridge install: YouTube video - details are in the video comments area - video was cropped, enlarged and slowed down. The video is unlisted so it won't show up in any Google search, but the link can be shared.

Highlights: Lots of manlift activity at the north end, the northeast (video front-left) corner of the span appears to make contact well before the northwest (video front-right) corner is set down, and at one point the north end was rapidly lowered - or dropped??

MikeW7 (Electrical)7 Jun 19 00:3

Mike, good observation.

I zoomed in some more. The northeast extreme corner of diaphragm rotated down and if it did not make concrete to concrete contact at pier surface, it at least went lower than shim top level. After that the corner had to raised back up to level diaphragm and clear shims. This video will fuel previous suspicion that span was rocked into place. Edit: Unfortunately, as I re watch video in loop, I see slightly different dynamics, so I'm not sure what I'm seeing. Also, trying not to blink puts me in trance.

Quote (MikeW7)

edited version of one of the FIU timelapse movies [...] slowed down

This is interesting for a variety of reasons. It really amplifies a defect that I now see is in the original videos: the frames are not uniformly spaced in time. All the downloaded videos (600x, 48000x, 1000000x) have this same issue. For example, in the 600x, which should provide the finest time granularity, they proceed frame, frame, frame, jump, frame, frame, frame, jump...

This suggests that in the original source video there's at least one more frame that was dropped, probably in order to result in a particular average frame rate.

Looking at video Bridge-109 Mar 8-19 2018 600X-1080.mp4, at the 14:41..42 range, it looks to me like the frame at which the bridge is first seen collapsed is right after such a jump. Which would mean there is a frame missing right before that (which could be just before collapse, just after, or during). That might be a very interesting frame.

qwideman - Yeah, the framerate on all the videos is dynamic. Something like a quick 3-shot burst then a long (5-10 second) delay. I initially thought a section of north video you mentioned had been edited out right before the bridge collapse (because of the way the lightpole shadows jumped like the second-hand on an analog clock) but finally decided the jerky gap was typical of the rest of the video.

I also examined the videos from the southwest and southeast cameras but there didn't appear to be any unusual movement when the span was seated.

One other oddity is that when the span was first lifted in the predawn (viewed from southwest) the span noticeably rocked side-to-side.

So much for insuring the span never exceeded 1/2 degree of flex/twist at any time. I wonder if the "computer shutdown" that caused the data loss was more than an accident...

One more reminder about Kevin Hanson - I'm sure he was present when the ends were tensioned before the move, and he was well aware of what the initial hairline cracks looked like. Immedediately after the span was seated (not sure if ends had been de-tensioned yet) he observed gaping cracks, took pictures (still unreleased) and forwarded them to his boss. It wan't until 3 days later that the other set of pictures (now publically released) were taken. The hairline cracks were indicative of an underlying defect, but the process of moving and seating the span is apparently what damaged it.

I created a couple of more videos from the north-view timelapse - one for the collapse day morning inspection crowd, and another for the bridge collapse starting when the mobile crane is fired up. All three videos are collected in a YouTube playlist. The videos and the playlist are all unlisted - the links can be shared, but they will not appear in a Google search.

EDIT - video comments are also disabled to limit my liability.

Quote (MikeW7)

One other oddity is that when the span was first lifted in the predawn (viewed from southwest) the span noticeably rocked side-to-side.

If you haven't seen it yet, there was an on-board video of the move, with the camera on the north end. This clearly shows the north end (at least, maybe it was the entire span) rocking from side to side, starting around 0.45 into the video. Link

And is there any chance a settling substructure could be one of the contributing factors? There is a lot of dredging being done in the canal before the collapse next to the newly-installed canal bulkhead on the south side of the canal. This was intended to support fill for the pylon (and the north end of the span), but I have yet to see any plans that show this bulkhead, as opposed to the bulkhead on the north side of the canal (for the adjacent span). The south side bulkhead was only needed once the bridge was shifted 11 feet north to accommodate a future road lane. Did they just duplicate the other bulkhead without any geotech studies or additional engineering? Also note that on the plans which can be found here Link, the canal bulkhead was designed to be only about as deep as the proposed bottom of the canal after dredging. Does this seem deep enough to protect the driven piles and thus support bridge piers?

Wetlander (Specifier/Regulator)7 Jun 19 12:11

The rocking is more apparent if you restrict your view to the distant horizon. The hydraulic lift system was not capable of lifting all rams simultaneously. Rocking is result of alternating pressure.

Please note in video what appears to be a cardboard box of thin stainless steel shims on top of pier.

I'll upload a bunch of videos tonight to a dedicated YouTube channel (unlisted). It will contain segments from the north and soutwest FIU timelapse videos, and a southeast timelapse by Peter Grant. As soon as the new channel is up I'll delete the old videos & playlist.

This video also captures eastward dip. https://youtu.be/R_1nogbOLQ0?t=90

Time to review this: https://www.nbcmiami.com/on-air/as-seen-on/Insuran...

Shocking.

P.S. Watch the ram action at 2 minutes, 10 seconds.

List of major documents, articles, etc:

FIU BT-904 Public Documents List

Transportation.gov 2013 TIGER Grant documents
TY Lin Design Criteria June 2014
TY Lin Design Criteria April 2015 revision (to move tower)

MCM-Figg Design Proposal (2015-09-30) - jasm (Civil/Environmental) 21 Apr 18 18:42
Figg construction plans signed by Denney Pate (2016-12-09) - jasm (Civil/Environmental) 21 Apr 18 02:37
Barnhart Crane & Rigging - Bridge Movement Monitoring Plan /with sensor locations (2018-02-13) - jrs_87 (Mechanical) 7 Jun 19 19:06
Figg Structural Analysis Presentation (2018-03-15) - Samwise753 (Structural) 7 May 19 01:22
FDOT Fact Sheet (preliminary) (2018-03-15)
Miami Herald comprehensive report (2018-06-14)
NBC6 Interactive Timeline (2019-03-14) - MikeW7 (Electrical) 30 Mar 19 17:15

FDOT Plans Review (date unknown) - SFCharlie (Computer) 30 Mar 19 18:01


Leave a comment if you want more links added.

This video suggests investigators found installed steel. https://www.nbcmiami.com/multimedia/WEB-3-MISSING-...

FDOT Plan Review (Attachment)

https://files.engineering.com/getfile.aspx?folder=...

Axial Capacity of Member 11
I don't think anyone has delved into the capacity of member 11 in any detail.
Here is how it compares to member 9 - these are quick numbers - refine them as you see fit.

                    Member 11              Member 9
Dimension                  24x21                24x21
Length (approx)           27 feet              20 feet         scaled from screen
Reinforcing               8 - #7 (7S11)        10 - #7 (7S09)  Sheet 8-98
load March 15(AM)         1600 kips est         816 kips est   (back of envelope calc)  

Does it strike anyone as unusual that 11 appears under reinforced if compared to 09?
With twice the load of member 9, 35% more length, and 80 % of the reinforcing of 9?
Does anyone have access to a quick design of these two columns/struts? Use 10% for eccentricity for now. I no longer have access to my RC38 manual, and I don't think it had 8500 psi concrete addressed.
The genesis of this may be in Detail A-A, sheet B-39, which has the note (members with no PT bars). That detail shows 10 - #7 bars. Details with PT bars seem to show 8 - #7 bars.
Temporary PT bars were installed in members 02 and 11 for transportation.
An opportunity for a wrong interpretation? I am not sure 10 bars would been enough - maybe so.
Thanks,

Bridge Movement Monitoring Plan (includes location of sensors) (Attachment)

There is a treasure trove of information regarding move.

• https://files.engineering.com/getfile.aspx?folder=ce6b6f34-b90d-4a71-a802-8

RE: Axial Capacity of Member 11
To pursue an answer to my own question, I found this :
https://courses.cit.cornell.edu/arch264/calculator...
which is a concrete column design/analysis program. It is limited to 5000 psi for concrete strength, however.
First off, if member 11 has only 8-#7 bars, it does not meet the minimum requirement of 1% reinforcing. I assume that is true for AASHTO also. This probably explains why member 09 has more reinf - with no PT bars, it is assumed to be in compression. With PT bars, it is assumed to be in tension, so minimum % reinf does not apply (?).
Checking member 11 I see loads of about 2050 kips design factored. I used 1.25 dead load, 1.35 live load, and a dead weight of 11 kips/ft and live load per foot of 90#X30 ft X 1.35, for a factored axial load of 2050 kips.
Using the Cornell program on-line and inputting 24"x21", 5 ksi conc, 60 ksi steel, and choice "B", Design of Column with Dimensions Assumed - it says we need 32.3 sq in reinf. The area provided by 8-#7 bars is 4.8 sq. in.
Using selection "A", Analysis of Column, same parameters, and inputting 12 - #10 bars, it shows a factored load capacity of 1556 kips.
I realize this is 5 ksi concrete and not 8.5 ksi, but that seems to be a lot more reinforcing than I see on the drawings.
Analyzing the section for 8 - #8 bars yields a capacity of only 1297 kips.
For these conditions the section is woefully under reinforced.
Can someone provide the capacity using 8.5 ksi concrete?
Finally, checking member 09 with 10-#7 bars, the program yields a factored load capacity of 1288 kips, which may work for member 09 when using 8.5 ksi concrete.
Comments?

RE: Axial Capacity of Member 11 Continued
In Section A-A, drawing B-39 I do not see ties wrapping/turning around each compression bar. Is this not required in AASHTO?
I see the bars 6" from the sides and the side bars much closer. In looking at the underside of 11, with the PT rod pulled out, I do not see any longitudinal bars and only a single outside tie with its bend maybe 3 inches past the bars at the corners if the detail is correct.

Member 11 was failing up its length.

Thank you,

Vance Wiley, since you are examining cracks, here is diagonal crack that appears to have traveled across diaphragm.

The device I marked here is one of 5 string pots Link used to record/plot vertical displacement (position) vs time.

Quote (Vance Wiley)

In Section A-A, drawing B-39 I do not see ties wrapping/turning around each compression bar. Is this not required in AASHTO?

It is for seismic zones. I don't think it is in FL.

I've pointed this out in the past, the odd detailing of the reinforcement. No longitudinal bars at corner bends?? The J-bars embedded in the slab are just floating there?? Maybe the shop drawings show something different.

And those cracks up 11 could have been following the PT bars, with the PT actively engaged in a last ditch effort at shear resistance.

Just a note (see attached file):

Tensioning #2 and #11 cracked #3 and #10 !?!? Did they have any control over this project?


FIU Pedestrian Bridge / BT-904 - Crack Inspection Report

Bolton Perez & Associates, serving as the CEI for the project performed a visual inspection of main span truss members on 02-06-18 after PT bars tendons No 2 and No 11 were stressed. Cracks were found at moment of the inspection. These have been identified per truss member and a consecutive number within the member. The intent is to monitor these cracks after the bridge is fully tensioned and the main span is at the final location.

The members showing these small cracks are truss members that share the same blister at the canopy of the already stressed members No 2 (stressed 1/30/18) & No 11 (stressed 1/29/18) . We believe, this first stressing operation has temporarily created tension on members No 3 & No 10; thus, creating cross sectional cracks transferring the tension loads to the steel on these members. No other truss members within span 1 show any cracks similar to these shown on members No 3 & No 10.

• https://files.engineering.com/getfile.aspx?folder=604b1f82-3987-4a54-8f64-0

A few other notes:

#11 failed early - See attached report. It was posted on the forum earlier and I only know it by its title "CRACKS_REPORT_AFTER_SHORING_REMOVAL", seemingly contradictory to other references as all being well after the shoring was removed. None the less, I refer to Photo #4 and #5 which show a longitudinal crack along #11 proximal toward the centerline of the member and from both sides. Below is a better copy of Photo #4, also taken from somewhere off of this forum. These photos were also revealed in the Preliminary Report as Photo #1 and #2.



I doubt that there is any reasonable explanation to disregard the the nature of this crack. Quite simply, #11 has been racked and will only get worse, and it did!

Also, I've taken some liberty with overlays to highlight the issue of vertical forces transmitted from #11 to the shims. They have to wend their way around both the horizontal drain sleeve and the four vertical sleeves. The 11/12/deck node needed all the help it could get and this didn't help.

• https://files.engineering.com/getfile.aspx?folder=c90cb869-6cee-45fb-b045-8

jrs_87 - Yes - that is a breakout forming on the east side of node 11/12. Your photo is remarkably clear and shows perfectly the 8" pipe sleeve thru member 12. That sleeve should have created the formed surface I expected to see in 12 at 20 inches or so below the top of the deck. In looking at the bottom of 12 hanging over the pier,I have not been able to see that formed surface. That suggests the bottom 2 feet or so of 12 is not still attached to member 12.
Have you seen a better photo which shows the sleeve formed opening?
Thanks for the reply.

Answer to Vance Wiley (Structural)8 Jun 19 04:13

These photos are clear. No sign of 8" pipe.

• https://files.engineering.com/getfile.aspx?folder=7923d8fa-eb61-4bf0-bcf1-c

jrs_87 - Very good photos - thank you.

Quote (Vance Wiley)

To pursue an answer to my own question, I found this :
https://courses.cit.cornell.edu/arch264/calculator...
which is a concrete column design/analysis program. It is limited to 5000 psi for concrete strength, however.

Isn't this based on ACI 318-08? Link
Where on P.137 of PDF we have this:

with phi = 0.65.

Have you been able to locate in the FIGG calcs how they sized the steel for Member 11? The 1% minimum is a baseline.

Looks like a classic diagonal shear crack. Imagine rotating Member 12 90-degrees to act as a beam, with the applied load being the horizontal component of force coming from diagonal Member 11.

TheGreenLama (Structural)8 Jun 19 11:57

Quote (TheGreenLama)

Imagine rotating Member...

I swear you read my mind. That occurred to me too yesterday, I'm very glad you affirmed it.

Some other random observations. Figg calculations seem to check if steel in only 12 can contain force from 11(?) We know 12 broke at deck cold joint from photos. We also can guess very base of 12 remained for small period of time as diaphragm slid down pier face since the top inside central edge of pier was chipped out. (Or burst force caused chip before deck moved) (Edit: PT bar more likely cause of chip) I cannot confirm, but In NTSB stills, I see what looks like shattered piece of 8" PVC pipe on deck near PT bar. If so, it got there from being scrapped off by pier.

My edited timelapse videos are finally up as a Playlist on the Whirled Gnus YouTube channel.The Playlist and videos are all unlisted so they can't be found by Google search. I didn't add any individual video descriptions - all the details are combined in the lengthy playlist Description. Asmuch as possible, the individual time frames are the identical - all "Move 2" videos start and stop at about the same time, etc. When I get a chance I'll see if I can add links to other relevant non-timelapse videos.

jrs_87 - Good catch. And if the 8" pipe was on the deck after the collapse, it was sheared or knocked loose on the way down the face of the pier. That supports the idea that the failure was at the surface of the deck, or mostly so.
It would be so informative if we could see a good photo of the top of the deck at the base of 11 and 12, all swept clean and all debris removed. That will probably be in the final report by NTSB. I think I saw they had cut out that section of deck.
I would like to see the condition of the 4 hoops of reinforcing across the cold joint - designated 7S01 .

Well, that didn't take long.... Eng-Tips lurker and YouTube conspiracist Construction Engineering, & EQUILIBRIUM is taking credit for the Whirled Gnus Collapse videos and using them to reinforce his old conspiracy theories and create new ones.

EDIT ADDS:
Went back and added creation details and source information to all video descriptions so the viewer knows how the videos were created, and where the original source video came from.

Reading some of the comments and CE&E's replies - this guy is a real piece of work: repeating disproven rumors as truth, creating outlandish new rumors, and flat out lying.

Quote (MikeW7 (Electrical))

Eng-Tips lurker and YouTube conspiracist Construction Engineering, & EQUILIBRIUM is taking credit

added:Thank you for adding the info to your videos!
Yes,
At first, I watched because he was showing photos that I was not aware of...
But then, I realized he could not read blueprints. He was dependant on his readers comments to set him straight.
He has had to change the name of his chanel at least three times.
He is unwilling to share his sources. (Many are seen here.)

TheGreenLama and jrs_87
The Cornell program IS ACI 318 - but it is the only one I found. I figured it would yield a number somewhere in the ballpark to give an indication of the capacity of 11. I certainly would not use it to design a column or critically judge a design.
I was unsure of the phi factor you provided also. And I guessed at load factors for self weight and live load. I should have done more research to make the findings more applicable to this structure. How much reinforcing would you consider requiring in 11?
(I hung my slide rule up 13 years ago, but one does not easily abandon a career of learning and curiosity.)
Also, I was working Thursday on the concept of rotating the structure 90 degrees to better illustrate the importance of node 11/12 to those who may not visualize the workings of a truss and how a joint can fail horizontally from vertical loads.
The load matching the shear in 11 to deck would be that of one end of an equivalent span of about 1160 kips/11 k/ft or 212 feet weighing the same as this structure per foot of length. I was planning to pose the question "Would you stand under something that big with cracks like that ?"
Great minds run together.
Thank you for your comments.

Possible cause of canopy damage to the 10-11 blister area:

• During the move the underside of the canopy south of 10-11 was anchored to the "arms" on top of the SPTM movers but the span bottom was resting on cribbing directly beneath the canopy anchor points. This allowed the canopy to move independant of the deck.
• The canopy at 12 was chained directly to the deck edges. The canopy could not sway sideways with respect to the deck.
• When the bridge was moved over the roadway median there were several instances where the south SPMT movers moved independent of the north movers, causing the span to spin and shift on the cribbing (watch the north end of the span).
• If the canopy flexed at all during this, the 10-11 blister area would be near the midpoint of section between fixed end at 12 and the flexed portion chained to the SPMT arms. The 10-11 blister appears to be one of the initial failure points.

Shortly after spin-and-shift events the move was paused and somebody was taken up in a manlift to have a look around. The moved resumed immediately after the manlift came down - before it was even parked.

Quote (MikeW7 (Electrical))

When the bridge was moved over the roadway median there were several instances where the south SPMT movers moved independent of the north movers, causing the span to spin and shift on the cribbing

The SPMTs axles all go up and down independently, you can see this in the ground level shots and time lapses. So going over the median "should" not of tipped the structure. We do see the structure tip when it is being jacked up and down, but it looks as if the trailing edge is raised first, and then the leading edge, so no twist "should" have been induced. Besides, the cracks were noted while the structure was still on the construction supports before the SPMTs came on site. So whatever broke was starting to break before the move. Could tensioning the PT rods in 2 and 11 have pushed the already weak design over the breaking point?

SFCharlie - I'm not talking about rocking or tipping the entire structure. The canopy is chained directly to the SPMT arms, but the deck is sitting loosely on top of cribbing. If the deck shifts sideways (as viewed in the Ground Level video) the canopy chains hold the canopy stationary with respect to the SPMT arms, so the truss area between the canopy and deck is being bent - the vertical line of "I" cross section is bending as the bottom line moves but the top line does not. Similarly When the deck spins on the SPMT timbers while the canopy remains stationary with respect to the SPMT, the truss area between the canopy and deck is being twisted - the bottom line of the "I" cross section is rotating about the vertical line, with respect to the fixed top line.

Based on the original SPMT locations, the north end of the canopy was probably supposed to be chained to some arms on top of the northern-most SPMT, but after the design change that SPMT was moved inward, and Barnhart ended up chaining the north end of the canopy directly to the deck edges.

MikeW7 - It looks like the cables run over the edge of the deck and attach to the transporter beams. I had originally thought they were connected to the deck surface.

Vance Wiley - Canopy chains attached to deck at span ends, but to SPMT arms at SPMT lift points:

Re: "Canopy Chains"

This is final plan for move, includes call out for canopy chains.

See jrs_87 (Mechanical)7 Jun 19 19:06 for link:

EDIT: I re-uploaded the images but they still look like crap. A "feature" of the forum I guess....

Some notes on the SENSOR instrumentation, taken from the Barnhart Movement Plan link provided by jrs_87 (Mechanical) 7 Jun 19 19:06


In the diagram below, columns 10 & 11 were not monitored at all, and rotation sensors were installed at the points marked DISP2, DISP3 and DISP4. These monitor the structure tilt in the lengthwise direction.



In the diagram below, at 5 equally spaced points (labeled 1-2-3-4-5, but the same as DISP1 to DISP5), 3 rotation sensors (labeled A-B-C) were installed at the deck edges and center. These sensors monitor the tilt of the structure across its width. NOTES:

• The canopy chain is shown on the Twist Angle cross-section diagram as being connected to the deck curb, but as explained in my next section, BRIDGE TWIST is measured at points 2 and 4, where the canopy is chained to the SPMT arms.
• The comment box at the bottom describes how they expect the previously observed cracks to behave.

In the Testing Notes below:

• STRUCTURE TILT is defined as the AVERAGE OF SEVERAL ROTATION SENSORS - I assume that will be DISP2-DISP3-DISP4 for the lengthwise tilt, and each set of A-B-C sensors at the cross section points 1-2-3-4-5 for the side-to-side tilt.
• BRIDGE TWIST is defined as the DIFFERENCE BETWEEN ROTATION SENSORS AT LIFT POINTS, which I assume are the A-B-C averages at cross section points 2 and 4.

My observations:

• Columns 10 and 11 are devoid of sensors. The sensor record will indicate "nothing happened" to them during the move.
• There are no sensors anywhere on the canopy, and BRIDGE TWIST is defined between two deck cross sections (EDIT: at "floor level") at points 2 and 4. The events I described in my earlier posts involve the canopy area at the 10-11 blister moving/flexing with respect to the deck, which is I assume remains rigid. The sensor record will also indicate "nothing happened" to this area.
• These two facts explain why the on-deck activity during the move "pause" did not seem to involve the area around columns 10-11.

It's been a long afternoon piecing this all together (and I'm hungry), but hopefully I didn't mis-interpretate anything from the Barnhart document.

One final note: Below is a drawing of the bridge as shown on the cover page of the Barnhart document. I can understand why it shows the exaggerated tilt, but why are the SPMTs not in the same location as they are shown in the sensor location image? They are shifted quite a bit to the north (right). Attention to details....

Green arrow show possible remnants of 8" PVC pile. While I believe it's unlikely these are actually remnants of drain pipe, I don't have any idea what they might be instead. Other markups apply to another post.

Still looks like 8" PVC pipe? See attached in next post for larger size.

Larger size image.

• https://files.engineering.com/getfile.aspx?folder=56a5cf33-2908-457a-abea-d

Quote (Vance Wiley (Structural))

a good photo of the top of the deck at the base of 11 and 12

The NTSB report will likely be out this fall. They will have a "Docket" on their website with the Evidence they considered.

MikeW7 re cables and chains bracing canopy during move - I like the chains better. Big chains well anchored. This seems like the place for something with very little stretch under load - so that means large steel areas. Cable is not called wire rope for no reason.
There was a photo of an early crack at the fillet of 11 to the deck, with a path on the east side being a projection of the lower side of 11 to the deck and the cold joint. Across the face of that fillet the crack traveled downward to the west face at an angle - kinda like a tension crack might develop under a lateral force to the west at the top - maybe from some wracking or wobbling during the move.
The NTSB report will be interesting - at least to this group.

Today's article from the station covering the FIU bridge trial reveals some FIGG text messages, although just one from Denney Pate's phone, as well as some of the differences between the Bolton Perez meeting minutes and the FIGG version of meeting minutes. https://www.nbcmiami.com/investigations/Judge-Dema...

Looks like the only thing Pate will be remembered for is his ugly pedestrian overpass and its collapse. What a way to end an otherwise remarkable career:

• Senior VP and principal engineer in an engineering company with offices in 7 states, involved in the construction of over 200 bridges in 42 states and 6 nations, with a total value of over $14 billion(US).
• P.E. in 16 states
• Designer of over 30 majestic cable-stayed bridges.
• An actual Hall of Fame member (see the video).
• A distinguished alumni of Auburn's College of Engineering.

If the phone has water damage then iPad reHab could be the solution. Or Louis Rossmann. Or any of the other dozens of phone rebuilders who will repair whatever chip(s) are no longer chipping. For this level of need, data recovery/phone rescue pricing is worth it. It is unlikely that any irreparable harm has been done, especially to the memory chips.

"How did your phone suffer water damage?"
"It fell in the toilet."
"Did you fish it out?"
"No, I couldn't."
"Why not?"
"Because it was in little bitty pieces."
"Why was it in little bitty pieces?"
"Because I'd just spent a half-hour working it over with a ball peen hammer."

My wife lost a phone into a toilet. Before she could decide on calling the building maintenance guys to get it, the auto-flusher sent it on its way. She has a larger phone now.

"Gator ate it...."

Louis Berger, will we ever see what your superstructure review consisted of?

Has everyone here seen this? https://www.youtube.com/watch?v=GMyvLeNOPOw

I think many here know about his invention and many successful cable stayed bridges. It's why many are puzzled that he was working on a post-tension concrete structure. It doesn't solve the question about why people familiar with post tensioned concrete were climbing onto a bridge that was spalling off 100 pound chunks from the most heavily loaded portion, especially when they knew the FIGG hadn't done any like that before.

3DDave 10 Jun 19 21:39

Not trying to be facetious here; my wife thinks they were hypnotized.

Quote (jrs_87)

I originally created this image at an early stage in this thread, and not knowing the dimensions, I drew the wireframe outlines of #11 and #12 far too wide. Their width should not encompass the two pairs of white vertical pipes/sleeves, a detail which is rather critical. (I realize this wasn't the main point of jrs_87's post, I just don't want to perpetuate an incorrect impression about this crucial area.)

A much better version of this idea was posted by

Quote (Meerkat 007)

24 Mar 18 23:47

Quote (Wetlander)

. https://www.nbcmiami.com/investigations/Judge-Dema...

Wow. This thing has really turned into a mess. It sounds like Figg is ready to file for bankruptcy, which might have limited their liability a little bit, unless they destroyed and altered evidence like the article suggests. Also, I think the first video clip is kind of hilarious, with FIU seeming to think that they can escape all future lawsuits by relinquishing a $5M claim against MCM. I'm sure they would love to get off that easy.

Brad Waybright

It's all okay as long as it's okay.

Things are heating up: https://www.miamiherald.com/news/local/community/m...

Note: At time of this post, article is being updated constantly by the Herald, is incomplete, and is poorly written.

Time required to install shoring ?
Does anyone have an idea of the time required to mobilize the transporters and get them under the structure so they could have prevented this collapse?
Where are they stationed - how fast can they travel, what set-up time is required - could they just have been driven back in a short time?
This might tell us when it became too late to do anything that would have saved the structure. Of course, there is the question of repairs - if one does not know why it is cracking, they do not know how to fix it.

Step 1 would have been to stop traffic under it, whether they could shore it up in time or not. And that could have been done in about 5 minutes.

Quote (Vance Wiley)

Does anyone have an idea of the time required to mobilize the transporters and get them under the structure

The SPMT's are assembled like an erector set from a bunch of smaller parts that each fit on a semi-sized flatbed. I reviewed the SW timelapse (Bridge-PG6-Mar 1-19-2018-1080.mp4) and it appears the first parts arrived on the 1st and they started the move the evening of the 9th. Most of the last two days were spent installing chains, instrumentation, etc. The move was completed the afternoon of the 10th, and the SPMTs were parked until the next day (11th) when the disassembly started. The very last load left the construction site a few hours before the collapse on the 15th.

I'm in the process of created videos for the multi-hour inspections that took place on the 13th and 14th. It wasn't just a case of a few people taking pictures for an hour or so - there were several crews at various times inspecting the deck (mostly at column 11-12 junction ), the underside of north pier-deck interface (by manlift), and walking on top of the north pier.

MikeW7 - Thanks.
That means basically a week to get shoring under the bridge - that is about the length of time the thing had sat unshored. The outcome was determined the minute it was set on the piers.
The progressive nature of the cracking developed over at least 2 days - maybe more. No crack monitoring established - no visual recognition of a failure in progress. And they all sat silent in the meeting.
Has NASA verified the existence of Black Holes of Intelligence? I'm thinking one developed at the March 15 meeting site.
I'll bet an attorney can find that they exist.

JStephen Step One - how true. And saved 6 lives and what - 12 injuries? And $42 million to date.

Vance Wiley - Kevin Hanson took his crack pictures within a few hours of the bridge being installed - you can watch his crew on top of the canopy in the Move 4 videos that started just as the SPMTs were removed: North view - SW view. My guess is that Hanson is the guy in the lime green shirt roaming the deck - the same shirt was observed on the guy wandering the deck on day of the collapse, and Hanson had to be dug out of rubble after the collapse, so he was probably the deck guy.

The reports coming out now said that Hanson was agitated/alarmed when he called his boss because of the gaping cracks that developed immediately when the rods were untensioned. The pictures he took (and sent to his boss) have never been released - court fight going on but last I heard a judge wants them released and added to the public record. In a story earlier this year his girlfriend said she didn't want him to go back on the bridge (he had to return the morning of the 14th - day before the collapse), I guess because of what he told her.

In the videos I've reviewed so far I didn't see anybody return to the bridge later that afternoon of the 10th or the next day (11th) but somebody was on the deck for a few frames on the 12th (in N view only - haven't looked at SW view yet). On the 13th and 14th there were crews on the deck on and off most of the day - will have videos up tonight, I hope.

Quote (MikeW7)

posts videos

That's some informative work you're doing. Any chance of adding some kind of date (or even better, date-time) overlay? That way we viewers could easily see how the inspections corresponded to other events in that critical week.

Also, you mentioned Hanson reporting cracks that developed immediately on untensioning. Do you have a source for that? Thanks.

Quote (gwideman)

Any chance of adding some kind of date (or even better, date-time) overlay?
...
Also, you mentioned Hanson reporting cracks that developed immediately on untensioning. Do you have a source for that? Thanks.

Tonight I plan on renaming videos so they start with Date, and in the video description I will add a time stamp at the end of the original video name, something like: video_name.mp4 starting at hhh.mmm.ss

I really have no way of adding a real-time time stamp to the video because they don't have an embedded clock.

FIU bridge worker took pictures of cracks before collapse | Miami Herald

‘It cracked like hell.’ Feds say engineers didn’t heed alarms before FIU bridge collapse

The Herald has re-written article: https://www.miamiherald.com/news/local/community/m...

Quote (MikeW7)

FIU bridge worker took pictures of cracks before collapse | Miami Herald

Thanks MikeW7. But...

"When Kevin Hanson noticed that the thin cracks veining a crucial connection in the Florida International University bridge had opened into gaping fissures, he pulled out his phone and snapped a few photographs. It was March 10, 2018, the day the prefabricated bridge had been raised over a busy commuter road."

So far as I noticed, that article does not say that Hanson's pics were taken immediately after untensioning specifically. It might well be the case, and seems plausible, just doesn't seem explicit in this article.

OSHA report released today (115 pages): https://www.osha.gov/doc/engineering/pdf/2019_r_03...

The amount of detail is extreme and perhaps not typical for OSHA.

gwideman - If you watch the Move 4 videos the canopy crews are working at the north and south blisters, assisted by a big yellow mobile crane. After the crane is removed, street traffic resumes. Pretty easy guess that those crews did the untensioning, but I may have been assuming Hanson took the pictures AFTER the untensioning. In any case, he took them on the 10th, as reported in the article, which is the day of the move. I doubt he would have allowed the untensioning if he noticed the cracks beforehand.

Pate has a BS and MS in Civil Engineering from Auburn. WTF kind of Engineering Department does Auburn have that would teach a student to respond with an answer like "the EOR acknowledged that his computations could not replicate the cracks and therefore, he did not know why the cracks were occurring." I hope they yank his "Outstanding Alumni" award and ban him from ever setting foot on the campus again. JFC....

Quote (MikeW7)

I may have been assuming...

This question seems moot now, as the OSHA report just released contains this:

"March 10, 2018 [...] As they [VSL] began to de-stress the PT bars of diagonal 11, cracks began to appear at multiple locations, most prominently at the construction joint of diagonal 11 and the deck and at the top of the diaphragm II. There were three VSL employees performing the de-stressing – Kevin Hanson (supervisor), Navarro Brown and Chester Ashley. Kevin is regarded as one of the most knowledgeable PT field personnel in South Florida. After observing the cracks, Kevin became visibly disturbed and informed other VSL employees of the situation. Kevin took pictures of the cracks, and sent them to his supervisor, Sam Nunez, stating that “it cracked like hell”, see Figure 23. Reproduced below is Kevin’s text. Ashley mentioned to OSHA that Kevin went to MCM to show the photos of the cracks. Sam Nunez of VSL stated during an OSHA interview that the photos he received subsequent to March 10 were different, showing spalling and cracks in the diaphragm II implying that additional cracks took place after March 10, 2018.
In a March 22, 2018 interview with OSHA, MCM superintendent, Ernesto Hernandez, stated that Kevin Hanson told Pedro Cortes (MCM in charge of quality control) that cracks were appearing at the bottom of diagonal 11 after de-stressing. Mr. Cortes examined the cracks and took pictures on March 10, 2018."

Also MikeW7, you may find the OSHA report useful in correlating your video efforts with the report's description of who the personnel were. For example, page 75: "March 15, 2018 Two structural engineers from FIGG, Denney Pate and Eddie Leon arrived at the site approximately 7:45 am to examine the cracks first by walking over the deck. Thereafter, they evaluated the cracks by using a man-lift for better access. Also present were MCM’s Rodrigo Isaza and Pedro Cortes, and BPA’s Jose Morales. Denney Pate and Pedro Cortes went up in a man-lift, and examined the cracks."

It appears to me grout was not given 28 days to cure before shoring was removed in casting yard. Is this typical?

Regrettably, the report misspells Denney as Denny twice.

Our conspiracy theory friend at youtube will be ecstatic his video was referred to on page 27.

The grout offers long term corrosion and freeze-cracking control by excluding water from filling the channels. It also provides some protection if, in the long term, the anchors fail. In the short term I see no benefit to the minor section change in the grout which, installed after post-tensioning, will not be compressed.

For those who have been following this story, and particularly the details of the #11-#12-deck area, these photos from the OSHA report finally give us a decent look at that area.
Fig 62:

Fig 63:


As a side note: Fig 63 says that #11's lower PT bar sheared. Yet I don't think that matches what we see in this picture:


...perhaps that lower PT bar was cut off on site in order to detach this large piece of evidence for transportation.

Fig 64


And for what it's worth, this should be "from north looking south", I believe.

There are numerous other photos of interest. It's well worth downloading the report and reading through it.

I'm disgusted by what I read in the OHSA. If there is no criminality here I will be shocked. The structure started sagging right out of the gate and progressively deteriorated with each ensuing step in the construction process. By the time the #11 PT rods were released, the deck had twisted itself away from the 11/12/deck node. They've known all along, we didn't. How did they not have basic information about sagging from the monitoring. There should be a database revealing the progressive distortion of this nature.

Quote (jrs_87)

Regrettably, the report misspells Denney as Denny twice.

I think the spelling is forever more going to be "Mud" after his quote about not knowing why the cracks were occuring.

When somebody mentioned that Figg was considering bankruptcy I couldn't figure out why because I figured they were big enough to absorb the financial repercussions, but when Figg's customers hear about Pate's "did not know why" and the company's entire "deer in the headlights" attitude during the days before the collapse they will bolt.

Quote (MikeW7)

(quotes me regarding Denny spelling)

Actually that was jrs_87

I agree the entire document is worth the download.
Page 52
apparent cracks north end of diaphragm

SF Charlie
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Quote (MikeW7)

when Figg's customers hear about Pate's "did not know why" and the company's entire "deer in the headlights" attitude during the days before the collapse they will bolt.

Aside from clearing up a lot of heretofore ambiguity about the initial failure location, what stands out for me about the OSHA report is that it doesn't reveal any new shop drawings showing extra steel, or FIGG calculations that might have been later and better and replaced the ones we've been looking at this last month.

So while earlier this month we discussed what appeared to be really basic and conspicuous design errors more or less in plain sight, and wondered whether that could be all there is to it, the OSHA report agrees that these analyses were in fact deficient in the ways we identified.

The fact that basic analysis and design were incorrectly performed, and not caught, despite obviously silly numbers (like no rebar required in the 11-12-deck connection) is not just a mistake on this bridge. It's indicative of FIGG's process, and casts doubt not only on their future prospects, but on all their other past projects that have not yet fallen down.

Quote:

gwideman - Fig 63:

At guy with tape measure - "Construction Joint" - -
WHERE ARE THE #7 HOOPS ??" There should be 4 - "Truss member 11 & 12 has 4 - ties "7S01" --
That is a clean joint - do you see any 1/4" intentional roughening?
No wonder it slid so easy.
And 11 did not "slide" all the way off cleanly. It clearly did take concrete from below the deck surface back in areas which were reinforced. But no reinforcing thru the "cold Joint" in Fig. 63.
But - how far from the face of the pier is the deck slab as it rested at the bottom of the pier? Did it break the extension of 12 off as it slid over the edge? There was an instant when the shear in the 24 x 10-1/2 extension for 12 was enormous.
EDIT: Thank you for the pictures - Big time!

Quote (Vance_Wiley)

#7 hoops [...] 1/4" intentional roughening

Yeah, I don't see those hoops either, but if they sheared they might be obscured under concrete dust in this picture.

As for the roughening, that lack of such is mentioned in the OSHA report as one of the problems. But there was so much wrong with that connection it's hard to know which part to be more exercised about.

When you look at Figg's portfolio the thing that stands out is the immense scale of their projects. I think what did them in is they really didn't understand how to downsize their engineering, in the way that a clockmaker skilled in making grandfather clocks would be stymied trying to make a watch mechanism. Us oldtimers remember what a disaster it was when the American automakers first tried to make smaller cars - they were clueless, for years and years and years.

I don't think Figg actively sought out this contract. The owner of MCM is an FIU grad, as is 40% of their staff. My bet is that FIU approached MCM and told them they had the deal, but on the condition they had to get Figg aboard (this was a first-time MCM-Figg collaboration) so they could design one of the cool cable-stayed bridges that was at the top of the FIU wish list.


But FIU also wanted the bridge built with ABC (to enhance their engineering school brand) so the result was a compromise of the worst kind, a faux cable-stayed bridge. That would explain why Figg refused to change the design when Tom Andres asked if they had considered building "a real cable-stayed bridge" instead of a "fake one."

In the NBC Interactive Timeline - dated 10:55 Am March 25, 2016 - Andres sent an email to a Figg engineer (Jonathan Van Hook) and asked
That would (sort of) satisfy the ABC requirement, but Figg apparently refused.

If you haven't heard the entire story of the bridge's history, it's contained in this immense Miami Herald article from 2018-06-14 in the section titled Bridge Beginnings.

I doubt that 11 took any material from the deck as much as the rebar that was part of 12 levered it out during the collapse. The prying action of 12 during the collapse would be much stronger than the loads applied by 11.

Some errors in OSHA report thus far:

P46, photos 4-6 refer to base of diagonal, not base of diaphragm.


P47, according to MikeW7 video, Move part 4 N view, PT in diagonal 11 was detensioned prior to 2.

Mike W77 - I worked for a company with a similar problem. A third party was to design the item, but our company was responsible for both the cost of manufacture and the ultimate performance. That third party was, politically, buddies with the CEO of the company that acquired us in a heavily leveraged buyout and didn't care how much money our group lost on this contract as long as we still rendered unto Caesar 10%+ growth year after year, leading to massive layoffs to make up for the losses. I think, because the original design was so flawed, that it more than doubled the original fixed-price contract jammed down our collective throats.

Please discontinue posting in this thread. Go to Part X thread815-454618: Miami Pedestrian Bridge, Part X for any further posts.

This topic is broken into multiple threads due to the long length and many images creating longer load times for some. If you are NEW to this discussion, please read the following threads prior to posting to avoid rehashing old discussions.

Part I
thread815-436595: Miami Pedestrian Bridge, Part I

Part II
thread815-436699: Miami Pedestrian Bridge, Part II

Part III
thread815-436802: Miami Pedestrian Bridge, Part III

Part IV
thread815-436924: Miami Pedestrian Bridge, Part IV

Part V
thread815-437029: Miami Pedestrian Bridge, Part V

Part VI
thread815-438451: Miami Pedestrian Bridge, Part VI

Part VII
thread815-438966: Miami Pedestrian Bridge, Part VII

Part VIII
thread815-440072: Miami Pedestrian Bridge, Part VIII

Part IX
thread815-451175: Miami Pedestrian Bridge, Part IX

Part X
thread815-454618: Miami Pedestrian Bridge, Part X

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