Did the contractor not submit steel shop drawings for your review? Did they show cjp welds and decide to do something else without telling anyone?

If you have space, one option may be to bolt on flange plates or the Kaiser bolted bracket.

Thx for your reply
Yes, All shop drawings have been submitted and signed by all stakeholders, also they have accepted their's fault but it is not solution, the main question is, what is the best solution?!
If we add bolts on flanges or any thing like that, we can not change the criteria that we need CJP groove weld in protected zone, in all connections we need cjp because of plastic hinge and high degree of rotation in this zone and high level of fracture in this location
So what we have to do?!

I'm not sure how practical this is but can they remove the welds and cjp? Is there a bevel on the wide flange beam as would have been shown on the steel shop drawings? (i.e. was beam ready for CJP weld?) Why was there not a special inspector for these welds? There appears to be a combination of errors going on here...

With that said, can you revise to a different prequalified connection? I had suggested the Kaiser bolted bracket. I believe that does not require welding. We've used it in the past to retrofit pre Northridge moment frames. Any potential for this?

The issue is not the connection but the built-up beam. Adding to the beam connection at the end will not address the built-up members deficiencies. The requirements for connections and members in AISC 358 are all tested, you can try and have your connection tested per chapter K. You are going to have to involve the authority having jurisdiction and setup a review panel. Have the connections tested and hope that the fillet welds pass, since you are testing you can also test a configuration with bigger fillet welds, etc

Dear jdgengineer You are true, there are a lot of issues that are hidden in this problem and another teams are checking legal documents, it has an expert welding inspector but ... :)
About CJP, comment of sandman21 is true completely,Connection Type will specify length of protected zone at face of column and in this zone we have to connect flange and web by CJP groove weld. Also in AISC 358 in built up section has been stated built up members should be similar to rolled section and i think it want to achieve a uniform section to see a uniform plastic deflections with minimum deficiencies and fractures. Therefore it seems the CJP is mandatory!
About Kaiser Connection in AISC 358 section 9.3.1 has mentioned Beams shall be rolled wide-flange or built-up I-shaped members conforming to the requirements of Section 2.3., then this type of connection needs CJP welds too.
Finally the only way is testing,is not?

So there is not any way except testing according to AISC-358 provisions ?
any thing like repairing or retrofitting ?
assume this building had built some years ago and web and flange of built up beams connected by fillet welds, there is not any way to retrofit them according to FEMA ?

I think that sandman's post was pretty spot on. If you wish to pursue an alternate compliance route, it's time to start talking to your AHJ.

In my opinion, the main lesson that the structural engineering community learned from Northridge is that structural engineers aren't great at predicting cyclic connection ductility. That's why decided to stick to systems that have been previously verified by testing (prequalified). Out of pragmatism, we do set the bar a bit lower for retrofits. It's a slippery slope to start classifying the correction of new build mistakes as retrofits however.

Is there any way to modify your system so that less ductility is required and some of the mure onerous requirements would no longer apply?

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

FEMA 351 addresses retrofitting as a reference from ASCE 41 but it doesn't look like it will help you too much here if you were allowed to use FEMA 351. I would have them gouge out the welds and install the connection like it was intended or look at another system if you qualify (like Sideplate), but I think that would still require removing what is in place so it doesn't interfere with the new system function.

Rantamental:
You don’t show any element sizes or thicknesses, nor do you spell out the beam depth (protected zone length). You do say 8 or 12mm fillets, each side, as a function of the web thickness; pretty typical built-up beam fab., actually/normally a function of the shear flow. It would be helpful if you showed a couple sketches of these beam connections with all dimensions and weld sizes, etc. The reason for the CJP weld, flange to web, in that region is the extreme energy absorption actively going on in that region during hinge formation and reversal, and the want/desire to be more sure that there is no unprotected weld roots, with lack of fusion, crack starters, etc., during that high stress activity. Could you back gouge, from one side only, through the fillet and web and into sound metal at the root of the other fillet, providing a new groove to be rewelded for length “d” (beam depth, plus) out into the beam. Half of these welds would be down hand and half over head, but from only one side. I would probably preheat (post heat?) the connection to slightly soften or relax the region re: residual welding stresses, etc. And, I would not over match (by much) the base metals with my weld filler metals and process, so as to allow as much ductility as possible. I’d sure want the advice of a good welding engineer/metallurgist in setting up my weld procedure, process, etc. Then, I’d probably take him with me to meet with the AHJ for their approval.

Sideplate still has the same CJP/fillet requirement A retrofit generally does not bring your building up to current code level, assumptions regarding the life of the building are incorporated this lower some requirements you would typically need to provide. FEMA 351 would require project specific testing to qualify your retrofit connection, so you are back to the same place you are at now.

Agree on 351, that's why I said "if you were allowed to use" which is highly unlikely. I wouldn't allow 41/351 etc.. rehabilitation methods on a newly built building if I was the building official or the owner.

I would ask AISC - https://www.aisc.org/steel-solutions-center/. If nothing else, they should be able to point you in the right direction.

Note: The OP is asking about welds for the built-up beams. I don't think he is asking about the beam to column connection welds.

Hi
Thanks a lot for your help on this issue, thanks to all
At first i want to explain just a little more about our project.
Our Structural system is special moment frames in two directions and general arrangement of frames is Tube in Tube system with a rigid core in middle.
In external tube , Columns are closely together and we have a column in each 3 meters, so the depth of beams will be limited about one-seventh of clear span of that, Assume that Longitudinal dimension of column is 600mm, So clear span of beam will be 300-60=240mm and total depth of girder will be limited to 240/7 ( Approximately 350 mm).
We have wide range of spans and wide range of girder depth too. they have been summarized in following tables :




Also Following detail is general type for our connections and beam splices.

We Asked this issue from AISC 358 committee and they offered 3 options
1- The contractor could remove and replace the non-complying members. I recognize that this would be very costly and possibly dangerous. However, this solution would bring you into compliance
2- The contractor could burn off the existing welds, properly prepare the webs for CJP welds, and then complete the CJP weld. I recognize that this would be very costly and possibly dangerous. However, this solution would bring you into compliance
3- Connections in IMF and SMF must be either prequalified or qualified by testing. Requirements are provided in Chapter K of the AISC Seismic Provisions. Though not exactly what was intended, you might choose to “post-qualify” the connections that exist in the structure. This would involve building test subassemblages reflecting the actual conditions and then testing them per Chapter K of the Seismic Provisions. If the tests are successful then the conditions can be left as they are. If not then they will have to be repaired. This option might be a good one for gamblers. The tests will not be cheap (over a decade ago I got an estimate of three tests required per configuration at about $50,000 per test). If the tests work the testing may be cheaper than the repairs. However if they do not work you will have the cost of the tests and the repairs

So it seems there is not any simple way !

Thanks for the updates. It's an interesting problem to be sure.

I've been thinking about how one might reinforce this if one were in fact alllowed to reinforce it. You know, first principles. And that got me to wondering: why are these demand critical welds?

At first, I thought that the answer would be obvious. You have to develop the plastic moment so you have to develop reliable horizontal shear transfer to go along with that. But that's not actually true. On the column side of the hinge, 100% of the required horizontal shear capacity is provided by the column panel zone. So there's no problem there.

With respect to important jobs that the welds are doing, all I can come up with is that they prevent the compression flange and web plate edge from buckling. And one would think that would be a relatively easy thing to reinforce for. But, then again, there is that clever engineer hubris problem that I mentioned earlier...

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Because the area at the hinge zone will see inelastic behavior, buckling of the web, force redistribution to the flanges as the web loses capacity, etc. Your analysis stops at what a beam would do under static elastic loading. Its possible that a thicker fillet weld or even current fillet would work but the demand on the connections and plastic hinges are such that testing is needed to understand the behavior.

Quote (sandman)

Your analysis stops at what a beam would do under static elastic loading.

Not so. My sketch and analysis specifically detailed forces associated with plasticity and the web and flange buckling that might occur under cyclic dynamic loading. One can only get so "dynamic" in a quick, 2D, stationary sketch. I get that testing is required for complete understanding and alluded to it twice above myself. I'm trying to dive a little deeper.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Your sketch is true at face of the column, also we check shear by plastic moment and value of this shear is clear, so we design dublar plate in column's web to control shear. When a girder play a moment frame role in cyclic loading it will see some rotations and deflections that did not consider in elastic design. Please watch cyclic test of wuf-w or RBS connections in youtube, the mechanism of wuf-w fracture is clear and plastic hinge occurs exactly in beam at face of column. The main issue is that we need a uniform section in this location to sure that it satisfies the criteria of codes for rotations (0.04 radian)
So the welding in this location is not according to capacities.
Therefore if beam fails before complete rotations beacuse of weakness in protected zone, ductility of system will reduce and whole of frames will collapse.

Quote (Rantamental)

When a girder play a moment frame role in cyclic loading it will see some rotations and deflections that did not consider in elastic design.

No worries, I'm with you on this. Everything in my previous responses assumed and implied cyclic, inelastic behavior.

Quote (Rantamental)

The main issue is that we need a uniform section in this location to sure that it satisfies the criteria of codes for rotations (0.04 radian)

Firstly, with the right reinforcement scheme, I suspect that you could create a section capable of achieving full plastic section moment over 0.04 radians. That's pretty much what I've been getting at.

Secondly, while I agree that it's necessary to achieve the 0.04 radians, I'm not sure that it's really necessary to achieve the full section plastic moment. All that matters is hitting that level of rotational ductility while maintaining a substantial measure of load resistance. Were the web flexural capacity to be considered sacrificial, you might get some degree of strain softening which is always undesirable. I would argue that the impact would be small and that normal strain hardening would kick in after an initial dip however.

Quote (Rantamental)

So the welding in this location is not according to capacities.

It's important to remember that, in these systems, we're primarily designing for a particular amount of rotational/displacement capacity, not a particular level of load resistance. Consequently, there really is no required load capacity for the welds in this situation other than that expected to arise as a result of rotation. The litmus test for successful performance is simply whether or not the welds remain in tact up to the 0.04 radian displacement so that the web and flanges continue to stabilize one another locally. At least that's my take on it.

Quote (Rantamental)

Therefore if beam fails before complete rotations beacuse of weakness in protected zone, ductility of system will reduce and whole of frames will collapse.

If you stabilized the flanges to prevent local buckling, I would argue that you likely would still achieve complete rotation capacity. You might even be able to achieve that rotation at a lower Mp as a result of the web not participating fully which would serve to increase the safety margins on the member of your system designed for capacity/over strength.

With this situation, I feel that the following is likely true:

1) You won't be able to negotiate any further with the AISC 358 committee. You'll simply have to do as they've instructed already. As such, I wouldn't waste too much time trying to argue your way into a lower cost solution.

2) My, very strong, gut feel is that you've got a great system here and that the code deficient welding really does not compromise that. Another thing that you've got going for you here is that the entire perimeter of your building is moment frame rather than just having a few isolated moment frames. And you've got the rigid core. That means a lot of redundancy so, if a handful of welds did give way AND that materially affected the performance in a few of those failures, would that really be such a big deal? I would argue not.

3) The cost of the remedy, per the AISC 358 committee, will be enormous. As the engineer of record, I would consider it my duty to the client, and society at large, to at least make a modest attempt at selling the AHJ on a non-tested reinforcement fix. That, even acknowledging that your odds of success would be extremely low.

4) The only way that you'll have a fighting change of being successful with #3 is if you're able to tell a good story about why the system is fine as is. A story that speaks to the fundamentals of the behavior that the code provisions are intended to encourage. I'm trying to help you come up with that good story.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Isn't it true that those welds became demand critical after the Northridge Earthquake revealed numerous brittle failures at this location in steel moment frame?

I know about Northrige failures in the ubiquitous CJP beam column connections. I wasn't aware of anecdotal examples of web plate to flange plate weld failures in built up moment frame beams. Certainly, I'd be interested to hear about such examples if they're kicking around out there.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

What you are describing is a welded flange bolted web moment connection, which is a pre-Northridge moment connection and was part of the reason for the failures. Northridge not only had weld failures but failures of panel zones and flange rupture, during a moderate earthquake. The engineer has an obligation to the client to provide the building they paid for, contractor errors of this magnitude should not be passed off by telling a good story. This is partially how we were caught "unaware" of the issues with moment frames before Norhtridge. We only have a small picture of what the building looks, removal of perimeter moment frames could have large impacts on the core structure which were not anticipated. You are going to have to come in compliance with AISC 358 or provide testing.

Quote (sandman)

What you are describing is a welded flange bolted web moment connection, which is a pre-Northridge moment connection and was part of the reason for the failures.

Not at all, assuming that you're addressing my proposals here. I'm talking about a system where the vertical shear resistance remains as it started: as passing through a web welded to the column flange. That ameliorates the differential vertical stiffness issues associated with bolted web connections.

Quote (sandman)

The engineer has an obligation to the client to provide the building they paid for

I don't know the contractural arrangements of the project. There are arrangements under which the owner may well share in the cost and schedule savings associated with avoiding a needless repair. Regardless, at the end of the day, all resources are society's resources and they should be deployed efficiently.

Quote (sandmam)

contractor errors of this magnitude should not be passed off by telling a good story.

Even if it's the correct story? You seem to imply that I'm being cavalier about the whole thing. I'm not. I'm attempting to be creative, flexible, and technically responsible. Everybody makes mistakes. And that includes engineers as well as contractors and fabricators. As much as possible, my preferebnce is to work with folks creatively and collaboratively to correct miststakes rather than to just beat them over the head with inflexible regulations. High seismic design, of course, has gone the other way out a combination of real and perceived necessity.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

You might even be able to achieve that rotation at a lower Mp as a result of the web not participating fully which would serve to increase the safety margins on the member of your system designed for capacity/over strength.

The system you are describing is the thought process for moment prior to to Northridge.

Quote (KootK)

Even if it's the correct story? You seem to imply that I'm being cavalier about the whole thing. I'm not. I'm attempting to be creative, flexible, and technically responsible. Everybody makes mistakes. And that includes engineers as well as contractors and fabricators. As much as possible, my preferebnce is to work with folks creatively and collaboratively to correct miststakes rather than to just beat them over the head with inflexible regulations. High seismic design, of course, has gone the other way out a combination of real and perceived necessity.

What part of your story is correct? You cant answer the fundamental questions regarding behavior or forces in the beam web during a seismic event or even testing. You would have to begin at that point to try and convince anyone that your creative approach satisfies the intent of the requirements. Your assumption regarding failures in perimeter moment frames not being a big deal, without knowing the configuration or performance criteria of building is also unconvincing to try and tell the story that the welds are satisfactory. Its not a matter of being cavalier its matter that your story lacks fundamentals.

Quote (KootK)

With respect to important jobs that the welds are doing, all I can come up with is that they prevent the compression flange and web plate edge from buckling. And one would think that would be a relatively easy thing to reinforce for. But, then again, there is that clever engineer hubris problem that I mentioned earlier...

Not sure how easy it would be to quantify the force required at this joint since the weld strains will need to be compatible with the distortions caused by local flange buckling that will occur after multiple cycles. Testing has shown that nominally sized fillet welds at this joint have proven to be the weak link (Link). The AISC 358 commentary mentions that most of the testing done on built-up beams used end-plate moment connections, and if we look at section 6.4 we see that the requirement for CJP welds is relaxed to fillet welds sized at 75% of the web thickness. I suspect that as long as the welds are sized to develop the tensile strength of the web (75% thickness), there is a strong chance that the system will behave as intended.

Rantamental...Testing will be required, no doubt about that. But I think you would put the odds of success in your favor if you tested a configuration that used fillet welds sized at 75% of the web thickness. The contractor would need to repair the welds that don't currently meet that, but that would be a lot less work than repairing with CJP welds. I would also make the welds demand critical given that yielding is expected in this region, although this is not required by AISC 341 as far as I can tell.

Quote (sandman21)

.The system you are describing is the thought process for moment prior to to Northridge.

I'd agree if my proposed shear resistance mechanism were soft or brittle. It's neither.

Quote (sadman21)

What part of your story is correct?

Possibly all of it. Definitely the significant observation that there is a limited demand for the web to flange welds to serve as horizontal shear transfer.

Quote (sandman21)

You cant answer the fundamental questions regarding behavior or forces in the beam web during a seismic event or even testing.

As I see it, I've been been meticulously answering all of these questions. What exactly is it that you believe that I've missed?

Quote (sandman21)

Your assumption regarding failures in perimeter moment frames not being a big deal, without knowing the configuration or performance criteria of building is also unconvincing to try and tell the story that the welds are satisfactory.

OP said that this was a perimeter tube moment frame system with tight column spacings (Fazlur Khan variety). I took him at his word, particularly after he posted details that would seem to confirm the assumption. The permiteter tube moment frame system embodies a lot of redundancy as I mentioned. It's an internet forum for Pete's sake. If we all hung back waiting for complete information, nothing would ever get done.

Quote (sandman21)

Its not a matter of being cavalier its matter that your story lacks fundamentals.

I disagree. If anything, I believe that I've been doing more than my share to chase down the fundamentals. Heck, I'm the only participant other than the OP to post a sketch. Most of what's materialized here seems to be just "follow the code" and the usual seismic catch phrases about ductility etc. I understood that OP wanted some out of the box problem solving and that's what I've been attempting.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (Deker)

Not sure how easy it would be to quantify the force required at this joint since the weld strains will need to be compatible with the distortions caused by local flange buckling that will occur after multiple cycles.

Probably not too easy. My intention was to try to hone in on just what we need the welds for and then, using that information, try to generate an intelligent reinforcement scheme. Similar to your recommendation regarding the 75% welds, one might just default to trying to replicate the original capacity/rstraint condition.

With reinforcement in play to address other issues, my biggest concern with my previous proposal would be the need for the potentially disconnected web to resist the moment implied by the eccentricity of shear delivery at the hinge (red stuff below). There's no doubt that the omission of the penetration weld make things less good. The more interesting question, in my opinion, is whether or not that makes things less than good enough. It's tough to say without testing of course.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

There's no doubt that the omission of the penetration weld make things less good. The more interesting question, in my opinion, is whether or not that makes things less than good enough. It's tough to say without testing of course.

Based on the research in the link I posted above, we know that nominally sized fillet welds are not good enough. For the built-up sections that were tested, the failures initiated with yielding, followed by flange local buckling, followed by flange-to-web weld rupture. It's clear that the distortional forces caused by flange buckling place excessive demands on the welds. Since these forces are next to impossible to quantify, the prudent course of action is to size the welds to develop the web. Although I appreciate your effort to think outside of the box, the analysis in your sketch could never have predicted that type of failure. In this case the decision to develop the web is not the default choice for the sake of conservatism, it's a carefully considered choice given all of the information available.

In my mind this is similar to the new requirement in AISC 341-16 to size gusset plate welds to develop the strength of the plate for SCBF systems detailed to buckle out-of-plane. Testing showed that weld tearing was caused by out-of-plane rotation demands on the gussets due to brace buckling. That never would have been captured by traditional analysis.

Quote (Deker)

Based on the research in the link I posted above, we know that nominally sized fillet welds are not good enough.

1) Any chance you'd want to reference some specific pages or sections to save me a bit of review time? I wan't to benefit from what you've shared but the document is 250 pages.

2) I mispoke. Please consider my previous comment revised as follows:

Quote (KootK ORIGINAL)

The more interesting question, in my opinion, is whether or not that makes things less than good enough.

Quote (KootK REVISED)

The more interesting question, in my opinion, is whether or not that makes things less than good enough when strategic reinforcing has been implemented.

Strategic reinforcing being something less onerous that gouging and full CJP of course.

Quote (Deker)

the analysis in your sketch could never have predicted that type of failure.

But I did predict that type of failure in my sketch as being the critical issue. That's what the squiggly line and the reference to buckling were about. And based on the research that you provided, it kinda sounds as though I might have been on the mark. So perhaps engineering judgement is still worth a little something after all.

Quote (Deker)

Since these forces are next to impossible to quantify

I don't see why it should be impossible. We do very much the same thing when providing restraint to rebar in the plastic hinge zones of shear walls etc. And again, a fall back position could be to reinstate the tensile capacity of the web by some means other than CJP welding.

I think that it's worth noting that we're all in agreement about most everything here. Very early on in this thread, I myself stated that 358 compliance would almost certainly be required. And I elaborated on why I thought that justified. I also brought up many of the same issues related to cyclic, inelastic behavior that are now being used to critique my own ideas. All I'm doing here is extending the conversation in an attempt to further understanding and creative problem solving.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

I'm going to continue to address any concerns raised about my work here. That said, I'm also going to try to scale back my involvement in this thread. I have empirical evidence that, when I start doing a lot of quoting, the villagers start to reach for their pitch forks.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Check out sections 6.5, 6.6, 7.2.4. When I read your statement I took it to mean that you could size the weld to prevent flange buckling. Based on the testing, flange buckling is inevitable and this creates a demand on the weld. That is the force that I believe is difficult to quantify. More of a deformation compatibility force, except the exact deformations are not known with any certainty.

Quote (KootK)

3) The cost of the remedy, per the AISC 358 committee, will be enormous. As the engineer of record, I would consider it my duty to the client, and society at large, to at least make a modest attempt at selling the AHJ on a non-tested reinforcement fix. That, even acknowledging that your odds of success would be extremely low.

This is only statement you've made that I take exception with. The testing has shown nominally sized welds to be deficient. I don't believe it is possible to prove otherwise without additional testing.

Thanks for the references. Certainly, I agree that an untested reinforcement fix should never be considered as rigorously proven as a successfully tested fix. I take that to be self-evident.

For interest's sake, imagine a hypothetical alternate universe in which Northridge had not happened and we were still at liberty to merrily apply our own, hubris laden engineering judgements. How would you go about specifying a reinforcement fix? I've shown a proposal of my own below for review. Some things that I was hoping to accomplish:

1) Prevent flanges from pulling away from webs.

2) Prevent reinforcement itself from increasing the plastic hinge yield moment.

3) Avoid unacceptable stress risers.

The one weakness that concerns me is that I worry that the pretension in the bolts might mess with the welds if the flanges to not bear tightly against the web.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

In the alternate universe I would still replace the welds with larger fillets since I believe that would be less work than what you propose...just a gut feel though. That said, I don't mind what you propose (in the alternate universe, of course).

Yes, the beefed up welds would be much less work. They'd also not encroach spatially which may well be critical. I'm confused though: do you think that the beefed up welds would work in Bizzaro world? I've been getting mixed massages on that.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

As I wrote in my first post, I suspect fillet welds that develop the strength of the web would work in the real world, but I wouldn't bet my license on it.

I don't have any experience designing special moment frames, but the challenge Rantamental is facing and the discussion in this thread has been quite interesting to follow. For my own education on this topic, I have been studying the AISC and NEHRP guidelines on WUF-W connections and wanted to raise some issues for discussion and to further my understanding.

1) The particular girder detail shown has 30mm thick flanges, which are slightly thicker than the 1" limitation in AISC-358. Also, the clear span-to-depth ratio (2400mm/350mm = 6.86) appears to be slightly lower than the minimum ratio of 7 allowed for SMF. In practice, are these slightly out-of-spec values acceptable for prequalification?

2) NERHP Tech. Brief No. 2 seems to discourage the use of deeper column sections: "deep wide flange
sections, particularly those with lighter weights, are susceptible to undesirable local and lateral-torsional
buckling. The performance of deep column sections is the subject of ongoing research". Wouldn't the skewed connection in this design exacerbate the likelihood of LTB? Can LTB be confidently resisted by lateral column bracing or is that one of the issues that future research will address? Is there any allowance for prequalified connections to be slightly skewed?

In a hypothetical alternate universe in which Northridge had not happened we would still have the requirements we have today. For one Kobe had similar failures in moment frames as Norhtridge, even if we assume that Kobe never occurred, the research for Northridge earthquakes connections showed these same failure modes in several tests during the 60/70. As more testing was conducted the requirements would have been developed.

As people have mentioned your first sketch is not a accurate approach when considering a moment frames behavior at .04rads. Large strains and forces develop in the interaction between the flange and web, which also increases the forces in the flanges. You are not the first to suggest beefing up the fillet welds as a means of retrofitting but that testing would be needed to determine the adequacy of the connection and the effects of the HAZ on the flange. A system which removes part of the web connection to the flanges exists SSDA, it still requires compliance with Section 2.3. Your new fix would have a large impact on the capacity of the flanges, likely causing a rupture to occur at the long slotted holes.

Quote (sandman21)

You are not the first to suggest beefing up the fillet welds as a means of retrofitting

I didn't suggest beefing up the weld. Perhaps you were addressing Deker there. I did find the parallels with the 75% provision in the bolted end plate section to be one of the more salient arguments presented here however.

Quote (sandman21)

A system which removes part of the web connection to the flanges exists SSDA

Facinating. Thanks for sharing that. Sans pre-qualification testing, if basically is what I sketched out conceptually.

Quote (sandman21)

Your new fix would have a large impact on the capacity of the flanges, likely causing a rupture to occur at the long slotted holes.

There already exists a pre-qualified bolted flange moment connection. Seems to me that the bolt holes should be resolvable.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (bones)

I don't have any experience designing special moment frames, but the challenge Rantamental is facing and the discussion in this thread has been quite interesting to follow.

I agree. I can't answer all of your questions but I'll chime in where I've got something to offer.

Quote (bones)

Wouldn't the skewed connection in this design exacerbate the likelihood of LTB?

I would think so. Firstly, your beam axis no longer runs through the column centroid. Secondly, there would be a component about the plastic hinge moment that would now induce weak axis bending in the column.

Quote (bones)

Can LTB be confidently resisted by lateral column bracing or is that one of the issues that future research will address?

You'll generally have two kinds of column bracing at each floor level:

1) Translational perpendicular to the frame as provided by infill beams etc.

2) Torsional provided by the weak axis bending of the frame beams and any supplementary bracing. Hopefully the frame beams are still capable of adequate torsional support once they go plastic. It's not a problem that's arisen in testing to my knowledge. It's an issue that would probably also benefit from the hinging not taking place right at the column flange face.

Between floors, the columns are usually unbraced of course unless you've got girts etc in play. That's fine so long as design assumptions reflect reality.

Quote (bones)

Is there any allowance for prequalified connections to be slightly skewed?

Since this got through plan check, presumably there is such an allowance. Perhaps OP can elaborate.

I was flipping through 358-16 last night to check out some of Deker's references. One thing that I noticed was that they've extended pre-qualification to some of the more exotic column shapes. Built up box columns, cruciform I-shapes, etc. Interestingly, the decision to include these shapes was based on judgment rather than testing. To paraphrase the reasoning:

Meh. So long as the inelastic hinging happens in the the beams, may the columns aren't such a big deal after all.

So I guess there's still engineering judgement at work on the west coast after all. It's just concentrated in the hands of the hyper-qualified/pre-qualified.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Quote (KootK)

I didn't suggest beefing up the weld. Perhaps you were addressing Deker there. I did find the parallels with the 75% provision in the bolted end plate section to be one of the more salient arguments presented here however.

You did suggest talking to the AHJ and trying to get a non-tested reinforcement accepted. One would assume you were talking about reinforcing of the welds. Regardless it has been suggested before to test larger fillet welds as a means to resolve the issue.

Quote (KootK)

Facinating. Thanks for sharing that. Sans pre-qualification testing, if basically is what I sketched out conceptually.

Your concept relieves no stress from the flanges and the slotted web still has the same CJP requirements other assemblies have. It would be a stretch to apply the basic concepts.

Quote (KootK)

There already exists a pre-qualified bolted flange moment connection. Seems to me that the bolt holes should be resolvable.

Are we in your hypothetical world or the current one? Does your hypothetical world have any pre-qualified? You cant go jumping from world to world without laying down some ground rules.

Quote (sandman)

Are we in your hypothetical world or the current one? Does your hypothetical world have any pre-qualified? You cant go jumping from world to world without laying down some ground rules.

Fair enough. The ground rules of KootK Hypo-World are these:

1) No regulations impede the free use of engineering judgment in any way.

2) All information from all worlds, past or present, is available to inform judgment.

In this way, one could know of 358 and the research that underpins it without being bound to follow its requirements.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

Dear Engineers
Thanks to all for your nice comments and beneficial negotiations
I have to say that we are consultant of this project and it is not possible for us to give or not a permission according to engineering judgment or theoretical sketch or traditional analysis! Any document should has a strong international reference and could be defend-able in any Court that may be formed in near future. I know, maybe (In real world maybe) that current fillet welds pass criteria of codes, but could you sign the permission documents and acceptance of current fillet welds with confidence with out any extra test? Also accept complete responsibility for this permission and defend it legally with out any testing ? Something about 1700 Tons of structure has been manufactured and erected with this deficit.

Quote (bones206)

1) The particular girder detail shown has 30mm thick flanges, which are slightly thicker than the 1" limitation in AISC-358. Also, the clear span-to-depth ratio (2400mm/350mm = 6.86) appears to be slightly lower than the minimum ratio of 7 allowed for SMF. In practice, are these slightly out-of-spec values acceptable for prequalification?

Yes, you are true but our documents for designing has a supplement that allow us to use built up beam with flange thickness up to 30mm instead of 25mm.
Also span to depth ratio is 6.66 in critical girder and ratio of allowable value to current one is (7/6.66=) 1.05 and we were allowed to going beyond up to this. (5 Percent in all sections)
Be careful that there are a lot of issues which mentioned in AISC358 sections and commentaries that are not clear in theoretical( Limitations: Unit weight of girders, height of beams, span to depth ratio, beam flange dimensions, value of Cpr and Etc) and adopted from a lot of tests. Now, Could we get a unique result from all discussions or not?

Rantamental:
We can NOT...
You get to pick what you can live with, prove to your satisfaction, and defend, we can’t/won’t do that for you. You certainly have been given a bunch of good, reasonably well thought out ideas and opinions, now it’s your job to apply your engineering judgement and experience to resolve the problem. You will have to be something of a diplomat to bring all the parties together, particularly the AHJ, and you hope they will be willing cooperative and constructive participants. When the fabricator, erector and GC consider the other possible alternatives, they should be able to find some money for testing, or whatever else is needed, to come to a reasonably acceptable resolution of this nasty situation.

Sure
I am with you and agree you
For us the most important issue is structure, safety and performance, so we offered just two following options
1- Test according to AISC 358 chapter k
2- change the members that do not comply criteria or change welds
It is clear that we can not accept any thing with out test, it is rule of engineering, try and error to find the way
Both of them are costly undoubtedly, but we could not find any better way, also I think the second one is impossible!!
Thanks to all for your helps and comments

One alternative that hasn't yet been broached here is performance based design. For a tall structure in which inelastic behaviour can be expected to be well distributed both laterally and vertically, I would expect joint rotation demand to be considerably less than the spec 0.04 radians. Even in high seismic regions, very tall buildings tend to be heavily influenced by wind performance concerns. The trick is that I don't know what, if any, lower value of joint rotation might obviate the need for testing. Perhaps other thread participants have insight to offer in that regard, even if it's to simply confirm that PBD is another dead end.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.