Moment Frame Fixed Base - Resolving moments into Couple Forces (Into Grade Beam)

[SoCal_Structural]

Hello Everyone,

I'm trying to check the connection strength from the wide flange column to the foundation (please see the link attached).

There is shear, uplift and moments applied to the connections. I know the connections are adequate for wind/seismic loads, however, this site is on a military base, therefore the buildings are designed for blast loads too. The blast loads that I've received are pretty high. I was hoping to resolve the moments (322 ft-kips) into tension in the (4) longitudinal bars that are welded to the base plate. However, resolving the moment into couples, yields a tension load of 215 kips. Since the (4) bars on its own aren't enough, I would have to count on the headed stud anchors in shear as well. The problem is, the headed stud anchors do not work in shear either. Basically neither the headed stud anchor nor the reinforcement are enough individually to resist the moments.

So my question is, Can I count on both HSA and the reinforcement simultaneously? If yes, how do I distribute the loads to the HSA as well as the reinforcement? Do I split the load 50-50 (or whatever the split maybe) and ensure the HSA and reinforcing is okay against half the total load demand? How would you go about this?

Thank you

 

[MrHershey]

One thing that may help is designs for blast loads are typically run at ultimate level, like a performance-based design. That means using expected material strengths (not minimum) and a phi factor of 1.0. A lot of times you'll also get to use a dynamic increase factor too for designs where actions have been resolved into static loads/reactions. All these in combination typically result in capacities that are maybe ~30%-60% higher than nominal, depending on material and what you're looking at. Would consult ASCE 59 or some of the USACE's Protective Design Center documents for more information.

As to the actual question, I have not seen anything regarding utilizing HSA and rebar in conjunction. Would be a little hesitant to just split 50-50 because the manner in which HSAs and rebar transmit forces to concrete is different. One is more of a mechanical anchorage and one is bond. Not saying you can't combine the two, I'm sure in the real world you can. Just haven't seen anything about what the distribution would be and don't know that it's prudent to just assume something because the mechanics are different.

 

[SoCal_Structural]

Thanks MrHershey! I understand the increased capacities due to blast loads. I already have the dynamic increase factors for both concrete and steel they allow between 15-35% increase in strength depending on loading case (shear/tension/bending). However, even with the increase in capacity, it's hard to resolve a moment of 322 ft-kips accompanied by a 39kip shear.

I'm thinking of recommending to widen the steel base plates to be able to weld more longitudinal reinforcing onto the plate, so that more rebars can be counted on to transfer loads into the grade beam. Either what I just said, or maybe allow for another safety factor, where I split the load 50-50 between the HSA anchors and rebars, but then increase my total load by another 25%. This would basically mean checking both HSA and rebars for 62.5% of the total load, which allows for some wiggle room given that HSA and rebars behave differently.

 

[sandman21]

If you know it works for wind/seismic, why are you not checking for the blast loading? The loads you are talking about are very low for high seismic areas and we easily get less reinforcing to work. My guess is you are using a very conservative method. AISC seismic design manual has an example of an embedded column, you should take a look at that. I would not use the HSA and reinforcing, you will have to check deformation capability between the two components and its not going to be 50/50.

 

[KootK]

I definitely would not consider the rebar and HSA in combination. Two reasons:

1) Strain incompatibility as MrH pointed out. You'll break out the HSA long before yielding the bars.

2) Equilibrium. The bars themselves have to deal with 100% of the moment transmitted to the the beam, right? If so, would not the added moment delivered by the HSA overload the beam in flexure?

Any chance you've not utilized your compression side bars in your calculations? I don't mean that to sound condescending. Just trying to hone in on the cause of the discrepancy.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[SoCal_Structural]

KootK, are you suggesting that the HSA anchors be removed altogether?

As long as I learn something new, I don't care. You are right, for the load transfer I think I've been way too conservative in counting on the tensile strength of the reinforcement only to transfer the load. Pardon my ignorance, so then I should go about finding the neutral axis and compression sides capacity (0.85f'c*Beta1)*c? And then add the tensile capacity Asfy to that? and compare against load demand?

I don't know why it didn't occur to me and I ended up just counting on tensile strength of the tension side reinforcement. Thanks again!

 

[sandman21]

HSA are added to transfer uplift forces from the frame to the gradebeam. Research has been conducted on using Wx beams as shear wall couple beams and this connection is basically the same. The design procedures found in Ch H of AISC 341 are based on this research and have been adapted for use in the condition.

 

[KootK]

Happy to help SoCal, truly.

KootK, are you suggesting that the HSA anchors be removed altogether?

No. As sandman mentioned, the studs are required for other, non-moment purposes. That said, I think that I see where your head's at. The studs aren't smart enough to know that they shouldn't be resisting moment no matter what you or I say. As such, will applied moment fail them and render them useless for tension? Maybe. Even if you break your anchorage frustum in flexure, you'll still have some -- surely reduced -- shear capacity as you attempt to shear that frustum through the remaining concrete under tension load. Nobody likes to admit to that but it's there, working in our favor at least partially. I've seen some beam/wall connections in the field where incidental moment at the connection produced a similar effect. Of course, there you've got infinite edge distance to work with. Head back in sand...

so then I should go about finding the neutral axis and compression sides capacity (0.85f'c*Beta1)*c? And then add the tensile capacity Asfy to that? and compare against load demand?

That's one way. There might be an even easier one. If your top and bottom beam steel are balanced, you may not even need the concrete at all other than for development & bar stabilization. Moment capacity would be [2 x As*Fy x distance between bar layers] with all the usual factoring etc.

Using, or at least bench marking, against the AISC 341 methods seem pretty smart to me. I don't know that the testing done on those connections considered the studs though.

My money says that, even without the studs, you'd have killer tension capacity simply by virtue of shear friction on your flexural bars and any concrete/steel compression induced during flexure.



I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[SoCal_Structural]

Exactly, like you said, there will be moment there regardless of me or you liking it or not. I wanted to mention that in my first post, but wanted to focus on the solution rather than trying to argue the point, lol. That was the first thing that triggered the domino effect of me over analyzing it. I find my supervisor evasive in instances like this (like you said nobody admits to it), but I'm glad to know that my thought process is somewhat rational even though it may not be conventional. Appreciate all the input everyone!

 

[KootK]

Well, your supervisor is your supervisor because she's good at keeping things moving forward and preserving fee. I do that too, just not here.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

 

[sandman21]

The HSA are not going to see flexural loads, the bearing capacity of the concrete along with the reinforcing welded to the column in a confined gradebeam will preclude the failure of the HSA. One large scale test of coupling beams without reinforcing. Large-Scale Testing and Analysis of Concrete Encased
Steel Coupling Beams under High Ductility Demands. If an engineer adds HSA's to the coupling beam they are used to control the slip, if any were to occur, which in your case is the uplift from the moment frame.