Continuous Beam Questions
Continuous Beam Questions
(OP)
What’s the most appropriate way to analyze a continuous steel wide flange beam over steel columns for flexure?
Consider a simple example:
-100 ft. long continuous steel wide flange beam (if there are any splices in the beam consider them CJP, effectively making it continuous. This may be unreasonable in practical applications, but it’s a needed simplification for me to ask my questions below)
-Beam supported on steel columns at 20 ft. O.C.
-Beam’s top flange braced by purlins every 5 ft.
-Beam’s bottom flange not braced by purlins
-Supports uniform gravity loads only
I’ve seen a lot of discussions on inflection points used at brace points in the past, but that AISC doesn’t allow that any more. What is the appropriate way to analyze the strength of this beam? To clarify, I’m not looking for advice on how to provide bracing to the bottom flange, I’m wondering what the proper way is to determine the strength of the beam as is.
Specifically, I have 2 questions:
1. What Cb value should be used? Do I use AISC 360-10 Equation F1-1, or some other equation? If I use F1-1, Do I use quarter points between the columns (as in, MA, MB, and MC would be 5ft. away from eachother), or something else?
2. What Lb value should be used? This seems to get overlooked in many of the discussions I’ve found on this subject. Taking Lb = 100 ft. seems way too unreasonable, the bending strength becomes so small that even a large Cb can’t make up for it. Taking Lb = 20 ft. for the column spacing might be reasonable, but I can’t find anything in the code that suggests you can do this. If you were to use the distance between inflection points for the negative moment you might get an Lb = 10ft.+/-, but this isn’t allowed by code.
To explain why Lb = 100 ft. for the full beam length is unreasonable, another way of looking at it is that if the beam were 200 ft. long (and Lb = 200 ft.) the bending strength would be even further reduced, but common sense tells us that a continuous 100 ft. beam over columns at 20 ft. O.C. should perform similarly to a 200 ft. continuous beam over columns at 20 ft. O.C.
Hopefully you understand where my confusion is coming from. Thanks for any help.
Consider a simple example:
-100 ft. long continuous steel wide flange beam (if there are any splices in the beam consider them CJP, effectively making it continuous. This may be unreasonable in practical applications, but it’s a needed simplification for me to ask my questions below)
-Beam supported on steel columns at 20 ft. O.C.
-Beam’s top flange braced by purlins every 5 ft.
-Beam’s bottom flange not braced by purlins
-Supports uniform gravity loads only
I’ve seen a lot of discussions on inflection points used at brace points in the past, but that AISC doesn’t allow that any more. What is the appropriate way to analyze the strength of this beam? To clarify, I’m not looking for advice on how to provide bracing to the bottom flange, I’m wondering what the proper way is to determine the strength of the beam as is.
Specifically, I have 2 questions:
1. What Cb value should be used? Do I use AISC 360-10 Equation F1-1, or some other equation? If I use F1-1, Do I use quarter points between the columns (as in, MA, MB, and MC would be 5ft. away from eachother), or something else?
2. What Lb value should be used? This seems to get overlooked in many of the discussions I’ve found on this subject. Taking Lb = 100 ft. seems way too unreasonable, the bending strength becomes so small that even a large Cb can’t make up for it. Taking Lb = 20 ft. for the column spacing might be reasonable, but I can’t find anything in the code that suggests you can do this. If you were to use the distance between inflection points for the negative moment you might get an Lb = 10ft.+/-, but this isn’t allowed by code.
To explain why Lb = 100 ft. for the full beam length is unreasonable, another way of looking at it is that if the beam were 200 ft. long (and Lb = 200 ft.) the bending strength would be even further reduced, but common sense tells us that a continuous 100 ft. beam over columns at 20 ft. O.C. should perform similarly to a 200 ft. continuous beam over columns at 20 ft. O.C.
Hopefully you understand where my confusion is coming from. Thanks for any help.






RE: Continuous Beam Questions
RE: Continuous Beam Questions
2. Lb would be the distance as defined in AISC "length between points that are either braced against lateral displacement of compression flange or braced against twist of the cross section." generally speaking in your case above for positive bending (compression in the top flange) Lb would be the 5ft purlin spacing and for negative bending (compression in the bottom flange) Lb would be your 20ft column spacing. You'll need to make sure you either have purlins at the columns or the beam to column connection and the column itself satisfy the bracing criteria of appendix 6 in the AISC specification.
To elaborate on 1 some more now that we have defined Lb, for positive bending you would calculate Cb for each 5 ft segment of beam and for negative bending for each column bay.
Patterning of roof live, snow, and rain loads will also have a large impact on the design.
Below is DL: 1 klf and LL: 1 klf on all spans with patterning considered:
RE: Continuous Beam Questions
RE: Continuous Beam Questions
Without some for of prevention for the lateral movement of the column, then LB would need to be considered the entire length of the beam.
At least that's just one guy's opinion.
RE: Continuous Beam Questions
Based on the envelope results to get, in my mind, an efficient beam design you'll most likely need bottom flange bracing at the columns and at the first 2 interior spans, well based solely on 1 klf once you start getting real loads those first two spans may end back up in positive bending.
RE: Continuous Beam Questions
Also - columns would typically brace the rotation via the column-beam attachment (bolts or welds) along with vertical stiffener plates from flange to flange.
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RE: Continuous Beam Questions
RE: Continuous Beam Questions
Dik
RE: Continuous Beam Questions
you describing the 5 span configuration in AISC table 3-22b in the steel manual?
RE: Continuous Beam Questions
2 beams 24' and 1 beam 52'... with end plate moment splices of approx 25 k-ft... and deflection of approx... 83/I (.00624ML^2/I for SS; M k-ft, L ft, I in^4). With continuity deflection is not normally an issue... loading assumed to be UDL. Done it so often, etched in grey matter...
Dik
RE: Continuous Beam Questions
RE: Continuous Beam Questions
Dik
RE: Continuous Beam Questions
As a hypothetical, say the purlins from the original example are smaller wide flange beams that run over the top of the beam in question (so there’s no question those purlins only contribute to bracing the top flange). Consider the column to beam connection is a top plate with 4 bolts into the bottom flange of the beam. The column could be wide flange or HSS. Would that 4 bolt plate connection prevent the bottom flange from rotating enough to effectively brace the bottom flange for negative moment?
Would it be reasonable to say it’s braced, or would you have to prove it via appendix 6? And if you use appendix 6, would the right check be ‘6.3 Beam Bracing – 1. Lateral Bracing’, or ‘6.3 Beam Bracing – 2. Torsional Bracing’? In other words, is the column preventing lateral movement of the beam’s bottom flange, or is it preventing rotation of the beam? This calculation would involve the strength and stiffness of the column as a whole and column height would even start to play a factor in this right? If I’m way off in my thinking just let me know, I’m not as familiar with appendix 6 as I should be.
Don’t get me wrong, at no point do I think you would have to take the entire 100 ft. length as the unbraced length, I imagine that would be way too conservative, even for the hypothetical above. I think an unbraced length of 20 ft. is reasonable, I just want to have a thorough understanding of the why.
Maybe some of this comes down to the fact that a continuous beam has moment reversals (double curvature). How can you assign an unbraced length to a compression flange for a length longer than that flange is even in compression for? I know… I’m going down a dangerous path towards using inflection points… but intuitively that seems to make sense. I suppose one of the dangers in using inflection points might be that, in this example, it could lead you to use an Lb = 10ft.+/-, but when considering live load patterning you could have a compression flange than remains in compression for even longer than the column span, such as 25 ft. long which would mean you’d have to use Lb = 25 ft.+/- by the same reasoning. You’d have to constantly change your Lb for every possible load combination and patterning, which would be very difficult to design for. Maybe you guys know of some other reasons why using inflection points is bad?
Sorry for the long post, I appreciate the discussion, let me know what you think, even if it’s just a response to part of this post.
RE: Continuous Beam Questions
I wouldn't.
I've seen numerous structural collapses in roofs where steel wide flanges were continuous over columns without any stiffeners.
I've never seen a collapse where stiffeners were present.
AISC deals with this by inserting the Cb factor that takes into account the curvature (reverse or otherwise) and applying that to the full column-to-column unbraced length to get you an accurate fix on your moment capacity.
Years ago I used to use the distance from the column to the inflection point - times 1.2 just as a feel good. I attended a seminar about that time put on by Joseph Yura of the Univ. of Texas.
A number of us asked him about this and after he thought a moment he said that he thought using that Lb (col - to - infl. point) was OK as long as you used Cb = 1.0.
One year later I was at another seminar and he had reversed his opinion and said you have to use the full distance with the proper Cb and that on no condition should you use the inflection point.
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RE: Continuous Beam Questions
No disrespect, but in MI we used it to reduce section size and facilitate the steel erection. No need for moment connections at the splices.
gjc
RE: Continuous Beam Questions
RE: Continuous Beam Questions
Dik
RE: Continuous Beam Questions
In case anyone is curious, I copied the OP *.jpg into Paint.Net program (a very good freeware graphics program) and exported it as a *.pdf. I then used Adobe free pdf reader and added the lines... Paint.Net is a little brother to Photoshop and extremely good.
Dik
RE: Continuous Beam Questions
The two shear-moment splices are bolted end plates and are cheap and easy to fabricate and erect... and there is a total of 3 members... Wanna bet what is cheaper... a whole bunch of Gerber connections and a whole bunch of Gerber members that have to be erected from several setups... and, compare the work for alternate loading and heavier sections to boot... a crane can easily handle the 52' member even with the 16' cantilevers hanging free in space (I'll try to dig up some photos of the long cantilevers) I'll keep doing it my way.
You'll have a tough time finding a section that is smaller...
Dik
RE: Continuous Beam Questions
I see your point for this 5-span situation. You'd have 3 pieces, 2 connections, and 3 crane positions. I'd have 5 pieces, 4 connections, and 3 crane positions. My connections would be shear only web plate connections and I would probably use slotted holes to accommodate any construction tolerances.
Cantilever pieces would be close to your sizes and hung pieces maybe could be slightly smaller. Probably not much overall cost differences.
gjc
RE: Continuous Beam Questions
Dik:
Is the plastic design covered in most steel texts or do you have a preferred reference for it? I have a very very vague memory of looking at this in college but really have not done much steel design in my career so far so very rusty.
RE: Continuous Beam Questions
Maybe partial height stiffeners would be appropriate in a situation where the purlins frame into the side of the beam; but for the hypothetical where purlins run over the top of the beam you would still want full height stiffeners. What do you think?
Edit: To clarify, I mean wide flange purlins with shear tab or double clip angle connections to the side of the beam. If it were OWSJ's I imagine it'd be similar to wide flange beams running over the top situation in which case you'd maybe want full height stiffeners (or some other way of bracing the bottom flange).
RE: Continuous Beam Questions
I just don't like bending webs from lateral forces.
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RE: Continuous Beam Questions
All my references had whiskers... they were that old. I had a couple of texts by Maissonette(sp?) and Save and also "The Steel Skeleton" by Baker and a couple of others... The first two volumes by M&S are extremely good... I don't know if you can still find them. I loaned the M&S to an engineer... that took off with them, never to be seen again, and, I gave the Baker book to a Mexican engineer about 40 years ago... There are likely good books on plastic design and most steel codes accommodate it and some concrete codes, often by allowing a moment redistribution. FYI, lower strength concrete behaves more plastically. About 45 years ago, I did a large warehouse for an engineer that had lost his license and used the Gerber system (only time I used it) and the engineer asked why I didn't use plastic design... read up on it over the weekend and have used it since... over the years have had numerous engineers comment that plastic design was 15% more expensive... but, any projects that I've had costed were 5% to 10% less expensive... I almost 'go out of my way' to use it.
RE: Continuous Beam Questions
I'm with JAE... even if not needed, if I have a beam going over a column, I use full height stiffeners. Personal quirk...
Dik
RE: Continuous Beam Questions
Prof Joe Yura did some research work on the subject of torsional brace stiffness requirements considering no, partial and full depth stiffeners.
Simply supported beam W12x14 wide flange, 24 foot span, midspan vertically-applied top flange loading, resulting in top flange in compression (bottom flange in tension) and he considered the following cases:
CASE B: TORSIONAL BRACE TO COMPRESSION FLANGE WITH ¾ DEPTH STIFFENER FROM TENSION FLANGE.
CASE C: TORSIONAL BRACE TO TENSION FLANGE WITH ½ DEPTH STIFFENER FROM TENSION FLANGE.
CASE D: TORSIONAL BRACE TO COMPRESSION FLANGE WITH ½ DEPTH STIFFENER FROM COMPRESSION FLANGE.
CASE E: TORSIONAL BRACE TO COMPRESSION FLANGE WITH ½ DEPTH STIFFENER AT CENTROID.
CASE F: TORSIONAL BRACE TO COMPRESSION FLANGE WITH ¾ DEPTH STIFFENER FROM COMPRESSION FLANGE.
CASE G: TORSIONAL BRACE TO COMPRESSION FLANGE WITH FULL DEPTH STIFFENER.
RE: Continuous Beam Questions
Where was the torsional brace applied? At the supports or at the point of load application.
Did he have a curve for the condition where the entire flange was laterally braced or a curve showing the full elastic moment capacity?
Dik
RE: Continuous Beam Questions
End supports are laterally and torsionally fully restrained, and variable torsional brace stiffness was at MIDSPAN.
'Ideal' brace stiffness was determined based upon achieving the critical load (approx 6.5 kips = the plateau on the graph) that represents a load level where the Lb = 12'.
RE: Continuous Beam Questions
RE: Continuous Beam Questions
The first thing they pointed me to was pg. 2-19 of the AISC Steel Construction Manual 14th Edition, there’s a section titled “Beams and Girders Framing Continuously over Columns”. It’s a very short section with lots of good figures, highly recommend giving it a read. Mostly in line with the information you’ve already provided.
Another thing they pointed me to was equation C-F1-5 in the commentary for calculating the Cb value for gravity loaded beams with the top flange laterally restrained. Until now I only knew of equation F1-1 for calculating Cb, I didn’t realize there were more equations for Cb in the commentary.
Some of you probably already knew all of this, but sharing just in case.