Use of Steel SMF in Wood Frame Building
Use of Steel SMF in Wood Frame Building
(OP)
I'm curious about how to brace the hinge points in a Reduced Beam Flange Special Moment Frame with a wood diaphragm. I've seen a lot of steel angle braces back to the wood diaphragm, but it doesn't seem like the wood diaphragm will be able to resist the code required brace load (6% of beam flange strength). This is especially a concern when the MF is installed at the front of the building, where there is diaphragm at one side only. What are some other details designers have used that can be shown to work for the code required brace load?
In most of our residential wood frame projects, the MF design is controlled by drift, and the actual member stress is very low, with D/C ratio of 0.25 to 0.4. This implies that if I apply the Omega level load, the MF will still remain elastic, and the hinge will not develop. If so, is there a rationale for reducing the brace load at the hinge locations, or simply not brace the hinge points, since the hinge will not form under Omega level load?
The above also lead to the question that if a SMF remains elastic, is it essentially behaving like an OMF with the built-in safety mechanisms? The drift calculation for SMF (Cd/R=0.688)is significantly lower than OMF (Cd/R=0.857). Is the SMF drift calculation accurate, if it is behaving like an OMF?
Thanks in advance,
Stan
In most of our residential wood frame projects, the MF design is controlled by drift, and the actual member stress is very low, with D/C ratio of 0.25 to 0.4. This implies that if I apply the Omega level load, the MF will still remain elastic, and the hinge will not develop. If so, is there a rationale for reducing the brace load at the hinge locations, or simply not brace the hinge points, since the hinge will not form under Omega level load?
The above also lead to the question that if a SMF remains elastic, is it essentially behaving like an OMF with the built-in safety mechanisms? The drift calculation for SMF (Cd/R=0.688)is significantly lower than OMF (Cd/R=0.857). Is the SMF drift calculation accurate, if it is behaving like an OMF?
Thanks in advance,
Stan






RE: Use of Steel SMF in Wood Frame Building
2) A detail that I would support would be a torsional bracing scheme using a short steel section with a bolted moment connection to the SMF beam and headered off to the primary wood framing tying in. This only works where your wood framing is perpendicular to your SMS beam.
I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.
RE: Use of Steel SMF in Wood Frame Building
I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.
RE: Use of Steel SMF in Wood Frame Building
Good luck meeting the stiffness requirement though. If someone knows how to meet the stiffness requirement (including effects of diaphragm deflection between other lateral elements) Id be interested to hear because I think it's nearly impossible with plywood diaphragm unless you have an unusually tight spacing of lateral elements in the perpendicular direction. A lot of engineers I know just outright ignore it. I don't follow that logic, I pretty much just use ordinary frames in wood frame buildings. Even in SDC D & E they are allowed with some exceptions.
If you really need a special frame I would consider using a premanifactured one such as Simpson Moment Frame or Hardy Moment Frame. Neither of these frames have bracing requirements along the beam span.
RE: Use of Steel SMF in Wood Frame Building
RE: Use of Steel SMF in Wood Frame Building
RE: Use of Steel SMF in Wood Frame Building
We've been getting a lot of push back from GC's about using Simpson SMF. The reasons being:
1. Simpson only allow 1/8" tolerance. This is a very tight tolerance for an existing building that is neither plumb nor level. Most GC's don't want to be responsible for a wrong measurement and end up paying for the frame themselves.
2. The GC's will not order the Simpson frames until the foundation is poured, and the anchor bolts are set. At that point it is a 6 week lead time. Most steel fabricators can fabricate SMF's quicker, and allow you to make adjustments in the shop drawing review process.
3. The GC's steel subcontractor will not install the Simpson frames, because they feel like they are losing business to Simpson. Residential GC's are not set up to be erecting steel themselves.
4. Cost. Most GC's tell me they can get RBS SMF's fabricated for less.
KootK,
Thanks for the detail. It makes sense. However, most of our SMF's in the Bay Area are parallel to the floor joists.
Anyone with any thoughts about the second half of my original post? I'm struggling to understand if I need to follow such onerous RBS bracing requirement by code, when the RBS in my case will remain elastic even under Omega level loading. I understand that the RBS will develop a hinge if R=1.0 (ground acceleration), but isn't the Omega load suppose to represent the maximum possible load the element will experience?
RE: Use of Steel SMF in Wood Frame Building
As mentioned previously it is my strong opinion you can not feasiby brace a SMF in a wood building. Therefore, if you are going with a conventional moment frame i would recommend cacling / detailing as ordinary.
RE: Use of Steel SMF in Wood Frame Building
2. There's no way the Simpson frames are more expensive than all the crazy detailing associated with SMF, plus designing for overstrength. You and your contractors are poorly informed. I can buy the tolerance thing but you have terrible GC's if that's the kinda stuff they're telling you.
3. No that's not what the overstrength factor is used for at all. It's to ensure certain parts of the building remain elastic during a seismic event so plastic hinges form where you expect them to (at reduced beam sections).
RE: Use of Steel SMF in Wood Frame Building
If you want to use the pre-qualified connection shown in AISC 358 you need to follow all requirements or get your connection tested without bracing like Simpson. See AISC 341-10 section E3.4b.
As a general note I always provide bracing at the plastic hinge even when the code allows for removal of bracing as the moment frame system behave better.
RE: Use of Steel SMF in Wood Frame Building
RE: Use of Steel SMF in Wood Frame Building
1) Overstrength loads are not viewed as the maximum load the connection can see. That used to be the case some years ago (1994 UBC-ish), but not anymore. At least not for SMF moment connections.
2) You've got options, OMF, R= 3 and such for certain seismic design categories.
3) Local buckling of the beam in the vicinity of the RBS is a real phenomena. And, it definitely reduces the ductility of the frame. You can't call yourself an SMF or expect an R value akin to an SMF without following that provision.
4) If you left out that bracing would your structure be "okay" under a seismic load? Maybe. Or, maybe you'd get some localized failure at the RBS that reduces the ductility. Maybe that impacts your structure as a whole. If you were dragged into court, then you would be responsible for not following the code provisions.
5) There are other outs allowed by code. But, they're expensive. You can go through 358 testing protocols for the type of configuration you want to use.
RE: Use of Steel SMF in Wood Frame Building
RE: Use of Steel SMF in Wood Frame Building
RE: Use of Steel SMF in Wood Frame Building
When elements in an RBS SMF building are designed to over strength, they are kept elastic by ensuring that they don't yield before the RBS's go completely plastic. The "overstrength" is the over strength of the SMF system. So, to check the RBS for over strength loading is to essentially say "I'm going to make sure that this beam remains elastic until well after it goes completely plastic!". Hopefully you see the incongruency inherent in that statement.
The only way to claim that the RBS's remain elastic is to design them for R=1.0. And even that's dubious.
I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.
RE: Use of Steel SMF in Wood Frame Building
There seems to be this continuing misconception about the overstrength factor. The overstrength factor is best described as the being the building relative overstrength. Placing overstrength onto any component does not ensure that it will remains elastic, it is elevating the loads to force your fuse to fail first and dissipate energy. Members that use overstrength could still experience inelastic behavior, they should not fail before the primary lateral elements dissipate lateral energy.
RE: Use of Steel SMF in Wood Frame Building
RE: Use of Steel SMF in Wood Frame Building
If it's a misconception, it's a misconception perpetuated by ASCE7. While the ASCE7 over strength provisions may not do a perfect job of keeping certain members elastic, that is ASCE7's intent. I see the Omega business as essentially a weaker, and often more convenient, form of true capacity design.
I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.
RE: Use of Steel SMF in Wood Frame Building
I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.
RE: Use of Steel SMF in Wood Frame Building
C12.4.3 description is the basis of my understanding of the Omega load. However, I know before Omega load was introduced, people used to use R=1 to check for elements that needs to remain elastic. My guess is that the code committee determined that R=1 is too conservative, and very unlikely for a structural element to experience that load level.
That's interesting that other codes use Cd=R. That makes sense if drift is calculated using ASD load. Seems like Cd should = 0.7R if using LRFD load to calculate drift. It'll be interesting to know how ASCE arrives at the Cd values. Maybe they are aiming for Cd = 0.7R with adjustments for different systems.
RE: Use of Steel SMF in Wood Frame Building
I can shed a little more light on this. At least... I think I can. In Canada, and many other countries, we have two R factors:
1) Ro (over-strength, 1/Omega) which accounts for the base shear differential between the point of first significant yield and the point where the entire structural lateral system is thought to have given way.
2) Rd (ductility) which accounts for the base shear differential between the spot that Ro gets you to and the R=1 demand level.
The sketch below does it better justice. In the US, Rd and Ro are combined into a single R factor. Anyhow, the moral of the story is that you only have to amplify by the Ro (Omega) to roughly ensure a member's elastic behavior. R=1 is indeed over conservative. I apologize if I misled you in that regard earlier.
I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.
RE: Use of Steel SMF in Wood Frame Building
That is my favorite diagram for explaining seismic concepts. It explains why we have R, Cd and Omega all in one simple force deflection curve.
RE: Use of Steel SMF in Wood Frame Building
In ASCE the Cd value is somewhere between Omega & R. Where Cd is closer to R value for more rigid systems such as walls, and Cd is closer to Omega for more flexible systems such as cantilever columns. Kind of make sense to me that in a rigid system, the element is more likely to experience ground acceleration, or R=1. If my logic is correct, the ASCE Cd value assumes that SMF (Cd/R=0.688) is more flexible than OMF (Cd/R=0.857). Which brings back my original concern that if the plastic hinges do not develop in SMF, then should we calculate the drift based on the SMF Cd & R values, especially when the SMF design is drift controlled.
RE: Use of Steel SMF in Wood Frame Building
RE: Use of Steel SMF in Wood Frame Building
The graph pertains specifically to US codes.
Nope.
I don't know. I'm not sure that anyone does really unless testing or at least a pushover analysis has been performed.
Personally, I think that drift should be calculated based on Cd = R for the reasons discussed above. Of course, that will have undesirable economic impacts in your marketplace.
Another wrinkle is that US R values have a bit of froo-froo fudge built in to pseudo account for anticipated redundancy. You can see this in the dual system numbers.
I observe the same trend but interpret it differently. With a flexible system, much of your R will be comprised of Ro (omega) and relatively less will be Rd(ductility). That make sense since a moment/cantilever frame will sustain relatively little damage at peak inelastic displacement. Thus Cd close to omega. With a wall, it's the reverse. Lots of damage at peak inelastic placement and thus a Cd value better correlated with Rd or R.
I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.
RE: Use of Steel SMF in Wood Frame Building
Thanks for the chart and explanations. I wish I had see it / read them years ago.
RE: Use of Steel SMF in Wood Frame Building
I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.