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Horizontal Trussing Threshold for Roof Diaphragm
3

Horizontal Trussing Threshold for Roof Diaphragm

Horizontal Trussing Threshold for Roof Diaphragm

(OP)
Hi,

How rigid must be the horizontal trussing in a roof before it can be considered a full fledge diaphragm? When the structure was constructed. They use 0.5x0.5 meter column from foundation all the way to third floor (it was designed for 4 storey).. but we decided now to build up to 3 storey only and plan to use thin metal roof to shed rain at gutter at side.

I know there must be a threshold in the horizontal trussing before the huge columns can become a diaphragm. For example. If you merely use a pole connecting the columns.. it can't be considered a diaphragm. Is it?

Also must the rafters be straight horizontal.. how about a bent gable rafter with apex at middle.. can a bent rafter create a diaphragm.. again what is the threshold of the rigidity? The pole is also rigid.. must it be certain strength compared to the huge columns?

Thank you.

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (mes7a)

I know there must be a threshold in the horizontal trussing before the huge columns can become a diaphragm.

I know of no such threshold.

Quote (mes7a)

Also must the rafters be straight horizontal.. how about a bent gable rafter with apex at middle.. can a bent rafter create a diaphragm

Non-horizontal rafters can support valid diaphragms. We do this all the time with residential wood trussed roofs.

I suspect that you would draw much better responses here if you posted some sketches of your situation. Is your diaphragm truly horizontal trussing with discrete members acting as truss webs? Or are you using the thin metal deck as your roof diaphragm?

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

I know of no such threshold.

The functions of beams or rafters as diaphragm is to equalize the loads between all the columns? But if the columns are huge.. the small rafters may not even move them.

Quote:

I suspect that you would draw much better responses here if you posted some sketches of your situation. Is your diaphragm truly horizontal trussing with discrete members acting as truss webs? Or are you using the thin metal deck as your roof diaphragm?

This is the layout.

..

The roofing is just thin metal or plastic sheet to shed rain with gutter at sides.. so there is no roof diaphragm.. just rafters diaphragm and they only rested at the columns (where the perimeter beam that holds the wall would be solely connected). What kind of diaphragm action would this produce?

The rafter is made of HSS that is sized 250mm x 100mm and 8mm thick. Only 3 pcs as the picture shows. The slope is as follows.

..

RE: Horizontal Trussing Threshold for Roof Diaphragm

What is the vertical lateral system here? Are there shear wall or cross braces? Or is the lateral system just the columns cantilevered up from the floor below?

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

What is the vertical lateral system here? Are there shear wall or cross braces? Or is the lateral system just the columns cantilevered up from the floor below?

No. It's purely special moment frames with no cross braces or shear wall at the transverse and longitudinal side. Yes, the lateral system are just the columns which started from the foundations all way to third floor.

The rafter is made of HSS that is sized 250mm x 100mm and 8mm thick. Only 3 pcs as the picture shows. The slope is as follows.

..

RE: Horizontal Trussing Threshold for Roof Diaphragm

You don't need a diaphragm here for equilibrium. You'd only need it if you felt that differential lateral column movement might damage the roofing.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

You don't need a diaphragm here for equilibrium. You'd only need it if you felt that differential lateral column movement might damage the roofing.

This is the whole building in 3D. Note the walls are only in the left and right side.. the middle front and back is open. The 9 columns are the sole lateral resisting system.

..

If I don't have a roof diaphragm. I'm afraid the floor diaphragm below may be stressed. What do you think about steel deck. Must the roof be made heavier to transfer half the wall load to the roof diaphragm? Without such.. I think it would go down to the floor below just you stated in some threads here.

RE: Horizontal Trussing Threshold for Roof Diaphragm

I stand by my previous post. Your walls can span up to your beams and your beams can span laterally over to your columns. If your metal deck had an established shear capacity, all the better.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

I stand by my previous post. Your walls can span up to your beams and your beams can span laterally over to your columns. If your metal deck had an established shear capacity, all the better.

What is the threshold to determine when the roofing would be cantilevered to the floor below and whether the roofing can serve as independent diaphragm.. when there is perimeter beams at top of the columns?

Also I'm reading a thread made in 2008:

http://www.eng-tips.com/viewthread.cfm?qid=219626

Msquared48 said:

"The "roof diaphragm" is the plywood membrane over the roof truss system, not the roof trusses themselves. Now the trusses have to be able to withstand wind and seismically induced loads applied in certain ways, but the structural diaphragm is the plywood. The roof trusses and intermediate eddge blocking serve to provide support to the diaphragm to stiffen it as necessary, but they are really not a part of the structural diaphragm per se."

So Kootk. Rafters are not considered diaphragm? But I read elsewhere where you said :

"Unless the top storey perimeter wall system is cantilevered from the floor below, which would be very rare, there pretty much has to be a roof diaphragm of some sort. That could include:

1) Roof deck acting as a diaphragm or;
2) Horizontal trussing in the roof plane acting as the diaphragm."

Msquared48 said roof truss system are not diaphragm.. you said it is. Can you please elaborate what is really the case?

I think diaphragms is to simply distribute the forces to the vertical lateral force resisting system. Why do the 2nd floor need slab diaphragm yet the roof doesn't need steel deck diaphragm? Unless you mean if the 2nd floor doesn't need to have slabs.. diaphragm are not necessary?

RE: Horizontal Trussing Threshold for Roof Diaphragm

You're making this a good deal more complicated than it needs to be mes7a. And you're clipping quote from threads that pertain to substantially different scenarios. Think of it this way:

1) If you don't have an equilibrium satisfying load path without a diaphragm, then you need one (usual case in steel).

2) If you do have an equilibrium load path without a diaphragm, then you don't need one (this case).

3) Even if you don't need a diaphragm, you may want one for the sake of structural efficiency (as with a concrete floor).

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

You're making this a good deal more complicated than it needs to be mes7a. And you're clipping quote from threads that pertain to substantially different scenarios. Think of it this way:

1) If you don't have an equilibrium satisfying load path without a diaphragm, then you need one (usual case in steel).

2) If you do have an equilibrium load path without a diaphragm, then you don't need one (this case).

3) Even if you don't need a diaphragm, you may want one for the sake of structural efficiency (as with a concrete floor).

Key is the word equilibrium. I can't find the word equilibrium on references on diaphragm, for instance. Diaphragm is defined in one source as:

"Diaphragms are horizontal elements that distribute seismic forces to vertical lateral force resisting elements. They also provide lateral support for
walls and parapets. Diaphragm forces are derived from the self weight of the diaphragm and the
weight of the elements and components that depend on the diaphragm for lateral support. Any roof, floor, or ceiling can participate in the distribution of lateral forces to vertical elements up to the limit of its strength.

In the above context. What do you mean by equilibrium? You simply mean to distribute the lateral forces? But why does steel not considered as equilibrium satisfying load path? Are you talking of steel column (noting my column is RC). Also noting my rafter is made of HSS steel. why does steel column not equilibrated compared to RC columns?

Thanks so much.

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (mes7a)

What do you mean by equilibrium? You simply mean to distribute the lateral forces?

Sort of. I mean a path that gets lateral loads from their point of origin down to the foundation.

Quote (mes7a)

why does steel column not equilibrated compared to RC columns?

Steel columns are often pin supported at diaphragm levels. Concrete columns almost never are.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

1. There is nothing unusual about using cantilevered concrete columns.
You just have to have horizontal members which deliver the load to the columns, and the columns need to be designed for the loads.

2. Steel roofing is to shed water, not to serve as a diaphragm.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

Sort of. I mean a path that gets lateral loads from their point of origin down to the foundation.

Ok. In your structures.. do you use steel deck as roof or thin metal or plastic roof to shed rain? Why would anyone use the heavier steel deck. They offer no resistances to falling airplanes too.. unless it is to be able to walk on the steel deck?

RE: Horizontal Trussing Threshold for Roof Diaphragm

When steel deck is used, the roofing is of a different type than metal (or plastic).

Falling airplanes? Is that a design condition for you?

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
Kootk, I need to know something. First. Have you ever designed building like the following where there is no shear wall nor is it braced frame.. but entirely special moment frame? not only that.. but all the walls are located at the sides. The middle is open (just glass at front and back).. meaning the heavy tranverse side is entirely resisted by the 9 columns during seismic lateral movement.

..

The building was originally designed for 4 storeys with concrete deck with all the seismic load combinations. But I lessen the actual building to just 3 storey with light roofing metal sheet.. I do this because you can't expect the actual construction to match every thing in the design or theoretical frame analysis.. therefore cut it one storey short to be on safe side (at least my philosophy).

The following is the column details for them. Only the 0.5x0.5 column is at center.

..

Now in the book Design of Concrete Structures (14th edition), there is this paragraph:

"The typical shape of a column interaction diagram shown in Fig. 8.10 has important design implications. In the range of tension failure, a reduction in axial load may produce failure for a given moment. In carrying out a frame analysis, the design must consider all combinations of loading that may occur, including that which would produce minimum axial load paired with a given moment (the specific load combinations are specified in ACI Code 8.10 and described in Section 12.3). Only that moment of compression that is certain to be present be used in calculating the capacity of a column subject to a given moment."

Now going back to the structure which is designed for 4 storey with concrete roof and only actually 3 storey built with light roofing. What is your experience about reducing actual storey? Have you ever done it? Generally, you may think the structure is even stronger with lesser floor actually built.. but in the tension part of the column, a reduction in axial load may not have the compression to make the tension part stronger (or below the balanced point in the interaction diagram). What do you make of this? What structure is sensitive to this?

RE: Horizontal Trussing Threshold for Roof Diaphragm

Since your columns have the same reinforcing pattern full height, I would not expect this to give you any trouble. When your building was four stories, you had a roof level condition where there was moment but little axial load. Now that your building is essentially three stories, you should have an almost identical condition at the third floor.

It's actually a fairly common occurrence for top floor columns to require the most reinforcing for exactly this reason. Out of curiosity, is the book that you referenced this one: Link? "Design of Concrete Structures" is a rather common textbook title. I guess structural engineers suck at product differentiation.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

Since your columns have the same reinforcing pattern full height, I would not expect this to give you any trouble. When your building was four stories, you had a roof level condition where there was moment but little axial load. Now that your building is essentially three stories, you should have an almost identical condition at the third floor.

But if the building is fully 4 storey. The tension part of the column of say the second storey would be stronger than if there is less axial load (from 3 storey). But then, for the compression side. Lesser axial load means more moment capacity (exactly the basis of the interaction diagram). You are saying it's more of a problem if the columns were tapered or made smaller as it reaches the roof? Since we use 12 to 14 meters bar length.. there is no splice from foundation to 3rd floor. The beams used were mostly 300 width by 500 depth, the beam bars were not spliced either because the building width is 12 meters side to side.

..

Just imagine the distance between the column is 6 meters each. Right now I'm trying to put walls and light roof to the third storey (presently it's just 2 storey with concrete roofdeck (same beams as below, designed for a floor). It's so hard to decide whether to use plain hollow block walls or lightweight precast.. or even just very light PIR insulation. I want the lightest wall to even have more safety allowance. Anyway the walls are just on left and right side.. no wall at front or back. This would focus wall seismic mass on transverse side. When the wall is on left side, and there is seismic movement to the right.. Can you say the floor diaphragm would restrain the wall seismic loads or would the columns mainly on the left restrain the walls lateral movement? And in special moment frame building without shear wall or not braced frames.. what is the componenents that make up the diaphgrams? slabs and beams?

I also am deciding between lightweight thin metal/plastic roof or heavier steel deck. But the latter seems much more expensive. However, a heavier steel deck would produce more axial load and perhaps strengthen the column-beam joint below it (?)

Quote:

It's actually a fairly common occurrence for top floor columns to require the most reinforcing for exactly this reason. Out of curiosity, is the book that you referenced this one: Link? "Design of Concrete Structures" is a rather common textbook title. I guess structural engineers suck at product differentiation.

Yes. It's that book where I memorized from cover to cover and my favorite structural references especially on manual computations.

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (mes7a)

You are saying it's more of a problem if the columns were tapered or made smaller as it reaches the roof?

Or, more likely, if the reinforcement was made smaller as the columns reached the roof.

Quote (mes7a)

Can you say the floor diaphragm would restrain the wall seismic loads or would the columns mainly on the left restrain the walls lateral movement?

All columns would share the load in proportion to their stiffness due to the presence of the diaphragm. As you mentioned above, one of the important diaphragm functions is load distribution.

Quote (mes7a)

a heavier steel deck would produce more axial load and perhaps strengthen the column-beam joint below it (?)

None of the roofing options that you've suggested would add enough compressive load to affect column flexural strength in a meaningful way.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

Or, more likely, if the reinforcement was made smaller as the columns reached the roof.

Why didn't you mention about the lower floor column tension strength. Perhaps you are saying that in the lower floor, the axial load is always on or above the balance point of the interaction diagram due to more axial load carrying the upper floors? But if the column is too massive like one meter across.. it is possible even all the load above would still make the lower floor column axial load below the balance point of the interaction diagram.. isn't it? For example. If you have only 2-storey building and your column is one meter in size. The axial load is below the balance point (of the interaction diagram) at the ground floor. What is the effect of this? Maybe lesser moment capacity but still enough for the ground floor??

Quote:

All columns would share the load in proportion to their stiffness due to the presence of the diaphragm. As you mentioned above, one of the important diaphragm functions is load distribution.

In moment frames.. many structural engineers design the perimeter columns as moment frame while the center is just gravity frame (do you also do this?). But in my structure, the center column is the strongest (being 0.5x0.5 meter in size compare to 0.5x0.4 meter at the sides). Do you think center columns can be a main lateral resisting system?

Quote:

None of the roofing options that you've suggested would add enough compressive load to affect column flexural strength in a meaningful way.

Ah. You mean it has only to be another floor to affect the lower floor column compression load in meaningful way. But I'm nervous following full 4-storey because you never know what the contractors do when you are not observing it. The following is the foundation plan of the building anyway (it's suggested by BARetired :) ) It's combined footing and we sized it 3 times larger to handle any trace of overturning moments etc.).

..

RE: Horizontal Trussing Threshold for Roof Diaphragm

I like your drawings mes7a. Very clean.

Quote (mes7a)

What is the effect of this? Maybe lesser moment capacity but still enough for the ground floor??

Exactly this. The textbook quote gives the impression that being below the balance point on the P-M is a bad thing. I think that it's the opposite. I'd rather be below the balanced point than above it. If your axial load is below the balanced point, additional moment will produce a ductile, flexural rebar yielding failure. The column will form a flexural hinge and cease to resist additional moment but it won't fail in the sense that it will lose it's axial capacity. On the other hand, if your axial load is above the balanced point, adding additional moment could result in a brittle, concrete crushing failure and the loss of axial capacity.

Quote (mes7a)

In moment frames.. many structural engineers design the perimeter columns as moment frame while the center is just gravity frame (do you also do this?). But in my structure, the center column is the strongest (being 0.5x0.5 meter in size compare to 0.5x0.4 meter at the sides). Do you think center columns can be a main lateral resisting system?

Any member in a building that would tend to strain under lateral load could potentially be used as part of your primary lateral system. For simplicity, we usually designate only some of these members to be part of our primary lateral system. In that case, the members not selected for primary lateral load resistance should be designed with sufficient deformation capacity that they can "ride along" with the designated lateral system without being compromised. I do this often.

Quote (mes7a)

Ah. You mean it has only to be another floor to affect the lower floor column compression load in meaningful way.

It may take 30 floors to have a meaningful effect. It depends on the size of the column.

Quote (mes7a)

It's combined footing and we sized it 3 times larger to handle any trace of overturning moments etc.).

Two issues:

1) This only benefits you in the left to right direction, right?
2) The footing/column connection at the end columns requires special detailing to make it work owing to the fact that the columns are right at the edge of the footing.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

Two issues:

1) This only benefits you in the left to right direction, right?
2) The footing/column connection at the end columns requires special detailing to make it work owing to the fact that the columns are right at the edge of the footing.

Do the following file pictures (of construction done 2 yrs ago) fulfill these special detailing? The following shows all the bars of the edge column put at the bottom bent... and the last picture shows the steel bars plans of the combined footing:

The following is before top bars put:

..

The following is after top bars and extra bars put:

..

The following is the rebars details of the combined footing (it's the top and bottom combined footing, the middle combined footing is large at 3 meters width and 12 meters length.. all has foundation depth of 0.6 meters 4000 psi concrete)

..

What do you think?

RE: Horizontal Trussing Threshold for Roof Diaphragm

You're detailing's pretty decent in my opinion. Better than most of what I see in practice. There's definitely a load path there. The only question is whether or not that load path can actually develop the required strength.

To be frank, I doubt that the connections that you've provided are adequate to supply the required strength (over-strength yield moment of the columns). Of course, I've no way to know for sure without running the numbers myself. That's just my gut feel.

Please don't interpret my comments here as mean spirited. My intent is to be helpful, not critical. One of the nice features of an anonymous forum like this is that we can discuss these kind of issues more honestly than we might be able to were we working in the same office.


I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

You're detailing's pretty decent in my opinion. Better than most of what I see in practice. There's definitely a load path there. The only question is whether or not that load path can actually develop the required strength.

To be frank, I doubt that the connections that you've provided are adequate to supply the required strength (over-strength yield moment of the columns). Of course, I've no way to know for sure without running the numbers myself. That's just my gut feel.

Please don't interpret my comments here as mean spirited. My intent is to be helpful, not critical. One of the nice features of an anonymous forum like this is that we can discuss these kind of issues more honestly than we might be able to were we working in the same office.

Does the required strength of the foundation-column joint depend on the soil below? Because we made the footing larger than it should be. This is because the soil report made an error. They reported silty sand and assume we would put the spread footing on the sand. But we dig half meter deeper and it is stiff tuff rock (we call it adobe) where we need to drill it to even touch the rock. Therefore the foundation should really be much smaller. But we already ordered all the bars and so we just build it. Would the soil or rock nature underneath affect what you stated above?

This is the front view of the top bars framing into the column (bars o.c. is every 6 inches):

..

This is inside the 0.6 meter depth combined foundation edge:

..

Anyway. My concern is the typical column-beam joint above in the 2nd floor. Many research are showing the ACI column-beam joint are not adequate. This is the reason that to be on the safe side.. I have to make it 3 storey actual when it is designed for 4 storey. Now my problem is the walls at the third floor. Right now the building is just 2 storey with roofdeck with one meter parapet. If I'd add 2 meters more of walls using 6" concrete hollow block. The weight of the additional wall at left side is 65 kN (2.73 kn/m^2 x 2 meters high x 12 meters span) and right side is also 65 kN. If we add the metal roof.. it would maybe reach 100 kN either side.

Now not confident of ACI column-beam joint detailing and we didn't use pure Vs in the beam shear reinforcement but Vc+Vs. I want to avoid more seismic mass. In your estimation.. do you think seismic mass of 70kN per side is large? Because if it is. I'd just use thin insulated panel wall that weights 20 times less for more seismic peace of mind. Many thanks.

RE: Horizontal Trussing Threshold for Roof Diaphragm

Wonderful photographs.

Quote (mes7a)

Does the required strength of the foundation-column joint depend on the soil below?

Not much. In a special moment frame, you have to develop a moment connection capable of resisting the flexural over strength moment in the column. I get that at about 270 kN*m.

Quote (mes7a)

do you think seismic mass of 70kN per side is large?

Nope. I think that it would be a drop in the bucket compared to the overall weight of the building.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

Not much. In a special moment frame, you have to develop a moment connection capable of resisting the flexural over strength moment in the column. I get that at about 270 kN*m.

Where did you get this 270kN*m? If the foundation column joint can't resist the flexural over strength in the column.. it would just break apart from shear failure? But if it's rock below.. the broken column would still bear on the rock, isn't it..

Quote:

Nope. I think that it would be a drop in the bucket compared to the overall weight of the building.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

But if you will see the following analysis of the dead load of the foundation using SAFE. It's 228 kN at the side. So 70 kN wall load would be significant. Maybe to be on the safe side. I'll just put insulated panel wall (lighter member sure can prevent more drift and hence make more resistant the foundation-column joint to break from shear failure. Isn't it?)

..

RE: Horizontal Trussing Threshold for Roof Diaphragm

I'm confused by the 70 kN business so I'll not comment on that any further. The moment demand on your column footing joint is a function of column capacity, not applied seismic load. 270 kNm was my estimate of that capacity at over strength. Soil bearing failure isn't the issue I was concerned about. The issue is the column separating from the main body of the footing.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

I'm confused by the 70 kN business so I'll not comment on that any further. The moment demand on your column footing joint is a function of column capacity, not applied seismic load. 270 kNm was my estimate of that capacity at over strength. Soil bearing failure isn't the issue I was concerned about. The issue is the column separating from the main body of the footing.

Ok. How would the maximum moment demand of the column footing joint occur occur? During seismic movement? Or are you talking of it outside the kern distance? What scenario would make it reach the maximum moment demand? Won't lessening the dead load help?

About the 70 kN. Please see the following sketch.

..

The weight of Concrete Hollow Block (CHB) is 2.73 kPa.. so a 2 meter additional CHB height (adding to existing parapet) would produce 5.46kN/meter. So with a span of 6 meters. it's going to be 32.76 kN (from 5.46 x 6). So the left side in picture above has total CHB weight of about 65 kN (or 32.76 kN x 2 beam span) that's the 70 kN estimate per side). You mean avoiding the 65 kN won't affect the column-foundation joint. But with 65 kN less. There is less base shear of the building. So it would take greater seismic movement to meet the moment demand of the column footing joint.. is it not??

RE: Horizontal Trussing Threshold for Roof Diaphragm

A special concrete moment frame has to develop plastic hinges at the column bases. As such, the moment demand at the base of the columns is equal to the over strength flexural capacity of the columns at minimum.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

A special concrete moment frame has to develop plastic hinges at the column bases. As such, the moment demand at the base of the columns is equal to the over strength flexural capacity of the columns at minimum.

I spent hours reading many references and books of how to compute for it. Do you know of a reference that shows how to compute for it. The Vc or shear capacity of the foundation at 2.5 meters width and 0.6 meters depth is about 1200kN. With 270kNm as moment capacity of column at over strength (at yielding plastic hinges). How do you compute for the moment connection capacity? I can't find it in ACI too. In the book it's written "To ensure the integrity of the junction between column and footing, ACI Code 15.8.2 requires that the minimum area of reinforcement that crosses the bearing surface (dowels or column bars) be 0.005 times the gross area of the supported column." For special moment frames.. any references exactly how to determine it? Thanks a lot.

RE: Horizontal Trussing Threshold for Roof Diaphragm

You can check the capacity using strut and tie procedures applied to the opening and closing models that I sketched above. Another way to look at it is to flip the model upside down in which case it becomes much like a roof level slab to column connection. The 270 kNm that I mentioned earlier was a ballpark number based on having to yield 8-20M column bars to over strength at a lever arm of 300 mm. I'd come up with a detailed estimate of your own before investing too much time on it.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

This thread has some relevant information on the design of column connections that transfer moment: Link

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

You can check the capacity using strut and tie procedures applied to the opening and closing models that I sketched above. Another way to look at it is to flip the model upside down in which case it becomes much like a roof level slab to column connection. The 270 kNm that I mentioned earlier was a ballpark number based on having to yield 8-20M column bars to over strength at a lever arm of 300 mm. I'd come up with a detailed estimate of your own before investing too much time on it.

Ok. Thanks. Btw.. as the following combined footing layout shows (posted earlier) and highlighted in blue. There is a tie-beam of size 300mm width and 400mm depth about 1.5 meters above the combined footing connecting all the columns together.

..

Here is the picture of one.. we first put compacted sand to fill up 1.5 meters of the combined footing then add a tie-beam connecting each column for moment redistribution effect and to make the columns stiffer.

..

Now in seismic movement.. where does the base plastic hinges form? below in the combined footing or above the tie-beam?

Also about 0.5 meter above the tie-beam are the ground floor slabs of 4inches in thickness. With all these stiffening and in your experience.. where would the plastic hinge form? in the ground slab level? In the tie-beams 0.5 meter below it.. or in the combined footing 1.5 meters below the tie-beam?

Many thanks for the tips.

RE: Horizontal Trussing Threshold for Roof Diaphragm

The tie beams will help a lot. You'd want the hinges to form at the top of the tie beams. You'd need to run some numbers to verify that though. That way, the end column to tie beam joint becomes critical rather than the column to footing joint. That's much better. Since your columns are bigger than your grade beams, developing the flexural over strength of your end columns may still be a challenge.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

The tie beams will help a lot. You'd want the hinges to form at the top of the tie beams. You'd need to run some numbers to verify that though. That way, the end column to tie beam joint becomes critical rather than the column to footing joint. That's much better. Since your columns are bigger than your grade beams, developing the flexural over strength of your end columns may still be a challenge.

In the edge, the connections are a big question mark. How about in the center column as the following bar details show (they are spaced 6 inches apart and 20mm grade 60). Do connections at centers better or look adequate to supply the required strength (over-strength yield moment of the center columns)?

bottom bars:

..

with top bars:

..

If it looks adequate. I am hoping the center columns of each combined footing can restrain the edge columns via the tie beams and perhaps create better resistances. Whatever. Since even the ACI doesn't have clear guideless on this. I'd just avoid greater seismic mass and base shear and just use lightweight panel walls for the third floor instead of thick concrete wall.. and also light roofing.. Thanks for this realization.

RE: Horizontal Trussing Threshold for Roof Diaphragm

The centre column connection looks fine. You seem to be missing a key point here however. Decreasing the seismic mass will not reduce flexural demand at these joints. Flexural demand is a function of supplied column strength.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

The centre column connection looks fine. You seem to be missing a key point here however. Decreasing the seismic mass will not reduce flexural demand at these joints. Flexural demand is a function of supplied column strength.

But it will take greater seismic magnitude to reach these flexural yielding demands if there is help from restraining parts like tie beam or lowering base shear.. right? so instead of magnitude 7 making it fail.. it would take magnitude 8 or sorta...

RE: Horizontal Trussing Threshold for Roof Diaphragm

We discussed that concept quite thoroughly in a previous thread I believe.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

We discussed that concept quite thoroughly in a previous thread I believe.

Yes. On beam rotations or plastic hinges.

About the foundation-column base thing. I think the analogy is the beam flexural strength compared to columns flexural strength.. the column flexural strength must be stronger than beam to avoid weak column-strong beam.. this is written in most special moment frames references.. but there is practically none on computing column base flexural strength to the footing connections.. if you encounter specifically them.. please let me know in future by giving the references on this thread...

I think this is what you mean by avoiding fixed base connections.. because you need advanced connections in the footing. We spent about $10,000 extra on the combined footing. But you know in the Philippines.. most footing are spread footing and they are eccentric.. ignoring the kern thing. Combined footing is so rare here. Originally, the width of 1.5 meter is enough.. but the designer has to make it 2.5 meters width to avoid overturning etc.. after I told him to design combine footing after understanding eccentric footings are problematic with the entire column not able to take the moments of the eccentric foundation.

I think not a single structural engineer in the Philippines knows how to manually compute the footing-column base yielding capacity. Do you know of any online service say in New Zealand or Australia where they can check structural plans from abroad and do manual computations of each with a fee. Our structural engineers forget how to manually compute. And the contractors don't have any ideas about basic principles. So we need international help if we want to scrutinize something.

Thanks for much for all the help. Structural enginners in my place know less than 30% of your knowledge. All they do is use software and that's all.

RE: Horizontal Trussing Threshold for Roof Diaphragm

I'd have to google firms in AU or NZ, same as you. In all honesty, concrete / seismic design and detailing seems a bit better in North America but not a lot better. And we're fast becoming software slaves as well. I wasn't being generous when I said that your footing detailing was better than most. It really is better than much of what is generated in my area.

Your building looks sturdy and well detailed in my estimation. I wouldn't be too hard on it or your design team.



I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

That sounds very frustrating. That scribbled word was "potential". And column / footing bar continuity is exactly the issue. If you google "concrete opening joint efficiency" you'll see what it's all about.

Your comment above about poorly detailed eccentric footings is interesting to me. I worked through one of those designs with an Iranian friend who said the exact same thing about eccentric footing design practice in Iran. It's curious why something so fundamental is being routinely misunderstood/ignored.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

That sounds very frustrating. That scribbled word was "potential". And column / footing bar continuity is exactly the issue. If you google "concrete opening joint efficiency" you'll see what it's all about.

Your comment above about poorly detailed eccentric footings is interesting to me. I worked through one of those designs with an Iranian friend who said the exact same thing about eccentric footing design practice in Iran. It's curious why something so fundamental is being routinely misunderstood/ignored.

The following is the original plan of the footing done by my designer. I posted it here 2 years ago to get feedback.. And many says it is silly, that the small column can't take the moments from the eccentric footing. So I got another designer with the idea of combined footing suggested here.

..

In the Philippines, most buildings have the footing above. The only reason for the combined footing thing is because of the feedback of people here. So thanks to them. Our designers mostly haven't done any combined footing. They just are good in eccentric spread footings.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
By the way Kootk... won't the ground floor slab contribute to any diaphragm action on the columns making the plastic moments develope on the slabs instead? The slab is 4 inches thick with 10mm rebar spaced at 300mm o.c. and the concrete used is ready mix with tested compressive strength reaching 5000 psi..

Are you saying that when the column bases reach Mpr (yielding plastic hinges), it can bend against the slabs and just flex the slabs or can the slabs somehow restrain the columns? What usually occur here? Thanks.

RE: Horizontal Trussing Threshold for Roof Diaphragm

Usually grade supported slabs are detailed with thin strip of compressible material around vertical elements like columns and walls. The purpose is to isolate these elements from one another for two reasons:

1) So that when the slab shrinks, the columns won't restrain it and cause unsightly slab cracking.
2) So that when the slab shrinks, it won't impose large lateral forces on the columns that could potentially fail them in shear.

Mostly it's #1. Anyhow, the end result is that your slab is generally not in direct contact with your columns and therefore is not a reliable form of restraint.

Even without the compressible strips, the slab on grade would still make for very questionable column restraint. If you worked out the compression force on the slab that would be delivered by the column in an over strength flexural yield scenario, it would be enormous. It would cause issues with:

1) The slab on grade crushing at the contact area.
2) The slab on grade buckling upwards.
3) The slab on grade sliding along the ground due to insufficient friction.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

A special concrete moment frame has to develop plastic hinges at the column bases. As such, the moment demand at the base of the columns is equal to the over strength flexural capacity of the columns at minimum.

But most special moment frame has pinned column bases (such as from spread footing).. you yourself said pinned bases are more normal because it's so expensive to make fixed bases (we spent over $10,000 *additional* for it and still not perfect).. the only reason for our fixed bases is because it is a combined footing..

For pinned bases.. can you still say special moment frame has to develope plastic hinges at the column bases? But if you will see the column moment diagram.. there is no (or not much) moment at the base because it's pinned so how can the base develope plastic hinges?

And why does special moment frame has to develope plastic hinges when the advantage of fixed bases are mostly in the ground floor (like tall ground storey).. and it won't affect the beam shear in the upper story nor the drift, as we discussed previously?

Anyway. If my foundation doesn't have the moment capacity to resist the column flexular over strength.. I could have use only a few bars (just like you use dowels) at beginning connecting to the bottom of the foundation to make it pinned, isn't it (obeying ACI 0.005 dowel steel ratio in the footing). But since the foundation already casted 2 years ago. Then this is is for knowledge and not actually trying to cut the bars to make it pinned.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
Kootk,

Quote:

The tie beams will help a lot. You'd want the hinges to form at the top of the tie beams. You'd need to run some numbers to verify that though. That way, the end column to tie beam joint becomes critical rather than the column to footing joint. That's much better. Since your columns are bigger than your grade beams, developing the flexural over strength of your end columns may still be a challenge.

The tie beams may not work. There is only 3 bars at the top of the beam (crossing joint) and 2 bars below. Remember the tie beam is 1.5 meters up the combined footing. Between it are soils. In my country, we use tie beam to restrain small eccentric footing. So the designer just left out the original tie beam to restrain small eccentric footing. He doesn't design it to make the plastic hinge develop above it. Anyway. I wonder how beam would behave if they are restrained by soil.. you can't call it a grade beam entirely because it is not rested on footing elevation. But then soil is soft material, so the entire column may just move against the tie beam and plastic hinges form back to the combined footing joint.

So I'll have to determine if the column joint right in the foundation has the overstrength capacity to handle the column moments. Before I'll search for any expert (if there is even one) in my country who can compute for it (or do you know of people right there in your place where they can do detailed calculations at a fee for peer review?). I'd like to understand some basic concept. I found the details about column at edge of slab at the book Reinforced Concrete: Mechanics and Design by James K. Wight. See:





There are many ACI methods to compute for it. However. It's mostly about slabs directly connected to columns above. Can you treat the foundation with edge column as slabs on columns and is the analysis the same? I'd want to explore the ACI methods. I don't want to try the harder strut and tie method you mentioned earlier because I don't understand them. I'm more familiar with the more familiar beam method or others that don't use strut and tie.

I need expert help. Do you know anyone in your place who can help? In my country. There is not enough sophistication to handle this analysis, although I will seek for one.. but need to know some basic so I can even ask the proper questions to him.

I want to determine if my edge column overstrength moment can be taken up by the foundation connection and detailing. If not. Then would go for retrofit of the existing tie-beam, maybe even replacing all the soil with concrete to change the plastic hinge above or reinforce the tie-beam with thick metal plates to transfer the plastic hinge above. Of course I'll find an expert in my country (if there is even one) for this. So I just need some initial thoughts of how to discuss with him.

Also if you will see the first picture above. The slabs are thin. But my foundation is very thick compared to the thin slabs presented above.. my foundation is 0.6 meters deep and 2.5 meters wide. Using the Vc equation of 2phi* sqrt (fc')* bw x d .. my Vc shear capacity is about 1200 kN. Now with 270 kN.m overstrength moment of the column.. can't you just use Shear x distance to get *rough* estimate of the moment capacity?

Don't worry. I'll let experts handle this. But just want initial guide of how to even present this to them. Thank you.

RE: Horizontal Trussing Threshold for Roof Diaphragm

You've taken some of my previous comments regarding pinned based columns out of context I'm afraid. I recommend pinned based columns for steel columns that are not a part of a moment frame. For special moment frames, I always utilize base fixity. This is because, in practice, it is very difficult to create a truly pinned column base and it isn't something that you would want to be wrong about in the case of a special moment frame where the stakes are high. And creating a truly pinned column base in concrete construction is particularly difficult.

Quote (mes7a)

And why does special moment frame has to develope plastic hinges

Because a special moment frame must form a complete mechanism at some point in its load history. Otherwise, it will continue to attract additional seismic load until some other, unintended member, reaches the limit of it's load carrying capacity and fails. This is a crucial part of the capacity design concept. Notice that, in the graphic below, yielding every beam at each end still would not produce a complete mechanism unless hinges also formed at the column bases. FYI: a complete mechanism is one which would render the frame incapable of resisting further lateral load. In a static load situation, it would represent collapse.

I think that the best bet for your building is to simply abandon the special moment frame concept. Analyse the structure as an ordinary or intermediate moment frame system where the demand on the joints is much less. I believe that this would work as your building is small and has several moment frames in each direction. Frankly, I very much doubt that it was necessary to eliminate the upper floor of the building. Do you have any concrete elevator or stair shafts in the building?

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (mes7a)

or do you know of people right there in your place where they can do detailed calculations at a fee for peer review?

The trick with this would be that, for most corporate entities, the risk/reward ratio would scare people off. I know that, if I brought this to my company, they'd be terrified of the liability and insurance issues and would charge you a ridiculous fee to compensate for that. That, or they'd just reject it outright. You'd probably need to find someone who possessed their own small firm and could do as they pleased. I can do one thing for you however. If you can talk your local engineer into letting you post his calculations here, I'll review them and provide comment. No charge of course.

Quote (mes7a)

Can you treat the foundation with edge column as slabs on columns and is the analysis the same?

Yes, quite right. I think that you might be gradually turning into your area's best structural engineer. The main reason for which I would prefer strut and tie is that such an analysis tends to do a better job of highlighting the need for developed top reinforcing close to the column. What it would amount to, more or less, is all of the moment at the connection needing to be dealt with by rebar that is:

a) crossing the potential yield line in the sketch that you clipped above and'
b) developed past that potential yield line.

Quote (mes7a)

The slabs are thin. But my foundation is very thick compared to the thin slabs presented above.. my foundation is 0.6 meters deep and 2.5 meters wide.

It's very astute of you to make this observation. The ACI slab method is intended for thin things and strut and tie is usually more appropriate for thick things. That being said, it's more about the ratio of column depth to slab depth than the absolute thicknesses. At 650/400, I'd still consider the slab method applicable. The Wight textbook is one of my favorites. I know for a fact that there's an excellent treatment of two way punching shear in there someplace.

Quote (mes7a)

Using the Vc equation of 2phi* sqrt (fc')* bw x d .. my Vc shear capacity is about 1200 kN. Now with 270 kN.m overstrength moment of the column.. can't you just use Shear x distance to get *rough* estimate of the moment capacity?

no. You're checking one way shear here when the primary issue will be two way shear. If you google those terms, you'll quickly get a sense for the difference.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

I actually checked punching shear on your column / footing connection back when we first started looking at it and it appeared that you had plenty of capacity there.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

I actually checked punching shear on your column / footing connection back when we first started looking at it and it appeared that you had plenty of capacity there.

Punching shear on the column/footing connection is not the same as flexural overstrength of the column footing connecting at yielding.. isn't it? Are you just referring to the axial load in the punching shear?

My designer with team of 10 all use software. They literally forget how to compute (because it would be taxing to manually compute 30 storeys). Last week I was asking for beam computations. They told me they can just output the results in Etabs. They design many 30-storeys in the country. See their website at:

http://www.rbsanchez.net/

They use SAFE software in the design of the foundation. SAFE should adequately handle punching shear.. but I don't know if it can check for design of column/footing flexural overstrength. I'll get the software myself to familiar with it so I can ask the designer next week how they designed it to check on the overstrength capacity (the person who input the program has already left their company so it would be others among them who would check it so I need to familiarize myself what to ask).

By the way. When the axial load is above the balance point of the interaction diagram, there is less moment capacity.. in other words.. can the column-footing really develop the overstrength at yielding without the compression side of the column crushing first?



RE: Horizontal Trussing Threshold for Roof Diaphragm

Thanks for sharing the link to your consultant's website. I agree, they have an impressive portfolio of completed work.

Quote (mes7a)

Punching shear on the column/footing connection is not the same as flexural overstrength of the column footing connecting at yielding.. isn't it? Are you just referring to the axial load in the punching shear?

The punching shear check accounts for both axial load and moment effects. As long as the applied moment in the check is that corresponding to the flexural over strength of the column, then flexural over-strength is accounted for. In my checks, I included the moment but ignored the axial. I did that because I don't know the axial and I suspected that the moment would dominate.

Quote (mes7a)

They use SAFE software in the design of the foundation. SAFE should adequately handle punching shear

As I mentioned above, it should just be a matter of applying the right moment to the safe model. Beyond that, there is nothing unusual about the check.

Quote (mes7a)

By the way. When the axial load is above the balance point of the interaction diagram, there is less moment capacity.. in other words.. can the column-footing really develop the overstrength at yielding without the compression side of the column crushing first?

This is a valid concern. It is for just this reason that the axial component of column stresses is kept low for special moment frames in ACI. They are limited to a small ratio of f'c I believe.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:


The punching shear check accounts for both axial load and moment effects. As long as the applied moment in the check is that corresponding to the flexural over strength of the column, then flexural over-strength is accounted for. In my checks, I included the moment but ignored the axial. I did that because I don't know the axial and I suspected that the moment would dominate.

you mentioned that column at edge of footing needs special detailing which involves top bars and if there is enough in in yield lines which needs close spacing. this is different from the punching shear you talked about above isnt it? because punching shear one way and two way shear can be calculated for footing with just bottom bar. so in the above description are you simply talking about the generic punching shear or did your calculations also include the yield lines of the top reinforcing bars of the flexural overstrength? if so', what computations methods did you use? many tnx. kootk.. you must be nominated as an officer of the Canadian Structural Society :)

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (mes7a)

so in the above description are you simply talking about the generic punching shear or did your calculations also include the yield lines of the top reinforcing bars of the flexural overstrength?

It was just the generic punching shear design and not the yield lines. There is an odd feature built into how punching shear is checked. Punching shear checks are predicated upon concurrent, correct flexural design. This is not always apparent to designers, however, as the presentation in codes rarely highlights this. Basically, if your joint flexural capacity is junk, so is your joint shear capacity. This article does a good job of explaining this: Link

Quote (mes7a)

if so', what computations methods did you use?

As I mentioned above, I would use the strut and tie approach for a high stakes situation like the connection between a special moment frame and its foundations. Using the slab punching shear concept is valid, I believe, and no doubt simpler to execute. I just don't have the same level of comfort regarding the results that I would with a strut and tie design. The sketches below show efficiency ratings for wall corner joints detailed in different ways. While none is exactly your scenario, I think that the sketches will help you to understand my concern for the careful detailing of the joint. If anything, the detailing would be more demanding in a column/slab joint than it would be in a wall/wall joint as the forces are more concentrated.




I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (mes7a)

kootk.. you must be nominated as an officer of the Canadian Structural Society :)

I wish. Such appointments usually go to those who are more skilled at self promotion than I. Thanks for the kind words though.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

It was just the generic punching shear design and not the yield lines. There is an odd feature built into how punching shear is checked. Punching shear checks are predicated upon concurrent, correct flexural design. This is not always apparent to designers, however, as the presentation in codes rarely highlights this. Basically, if your joint flexural capacity is junk, so is your joint shear capacity. This article does a good job of explaining this: Link

Have you used the software SAFE? Does it include the yield lines in the determination of the punching shear design? If manual would show it is inadequate.. then retrofit solutions would be logical.

In many designs of eccentric spread footing in my country. They put tie beam to prevent settlement and to distribute the moments. It is not a true grade beam to create base fixity.. which you mentioned before needs to be deeper and to be at the elevation of the footing surface. So the above tie beams are just standard we use.. not designed to transfer the plastic hinges above. But imagine the column-tie beam joints are being pinned from lateral movement from fixed.. in normal column-beam joints, the deflections occur because the beam is free to deflect.. but in the picture above.. the compacted soil restrain the beam from deflecting below.. would this increase the flexural capacity of the tie beam? Again the tie beam has only 3 pcs of 20mm rebar at top of its joint (for positive moment) and 2 pcs below. This won't make it great flexurally.

By the way. About eccentric spread footing. It doesn't have top bars. If the joint doesn't have the overstrength flexural capacity.. would it have the same punching shear capacity deficiently or would it compensate by becoming pinned when the entire foundation just rotate against the soil?

Going back to retrofit. Straightening the tie beam joints by enclosing it with metal plates or putting pedestral or even putting mini columns in between the tie beams to support the tie beams may make the connections more fixed and transfer the plastic hinge above.. isn't it.. what is your experience in this and what could be possible retrofit you think can work in the column-tie beams joints and below to make the plastic hinge develop above? Many thanks.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
By the way Kootk.. is the potential yield lines where the cracks would be?



Or is it in the same area as the 2 way punching shear?



Do you have picture of any such cracks where it differs from the punching shear cracks?

Also in sides.. the columns are bowing inwards because of eccentric loading so the CLOSING case is more the average situation as follows right? so the very closed spacing of longitudinal bars near the yield lines is not the for the "closing" loading case or scenerio.



I think the opening would only occur during reverse cyclic loading.. but if the columns are bowing inwards in the sides of the structure.. won't it be possible there is no reversed cyclic loading and the "Opening" case won't occur?

I need to present this to my designer. Unfortunately he doesnt understand the meaning of stress-strain diagrams last time. Because last time he told me to inject epoxy to a 2 inch void in the columns.. but hokie adviced against it saying the epoxy modulus of elasticity differs so much to concrete that it won't be compatible in the strains and compressions. This is the reason I started to focus understanding manual computations of the interaction diagram and the structure to understand the loadings.

Many many thanks. I want to send you amazon gift coupon if possible for giving so much useful tips :)

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

It was just the generic punching shear design and not the yield lines. There is an odd feature built into how punching shear is checked. Punching shear checks are predicated upon concurrent, correct flexural design. This is not always apparent to designers, however, as the presentation in codes rarely highlights this. Basically, if your joint flexural capacity is junk, so is your joint shear capacity. This article does a good job of explaining this: Link

I have read your link over and over again as well as more references for hours in preparation to presentation to the designers. Last time they failed to understand what I was explaining about stress-strain diagram and pure epoxy (not epoxy grout) filling up of voids. In our country. We fill up even 2 meters of voids in high rise building column bases with epoxy (there is no other solution).

Anyway. In reading all the references about yielding lines at slab-column that doesn't use beams. I realize they are all about your CLOSING case.. because the slabs don't anti-gravitate up the air... it goes down or closes with respect to the column. There in the foundation analogy, the bottom bars are the concerns. The yield lines in the books concern the bottom bars. My bottom bars near edge were extensive.

The references on slab-column without beams doesn't handle your OPENING case. Right? There if the top bars of the foundation near the edge is insufficient.. there is possibility of punching shear failure at column flexural overstrengh. However if the punching shear capacity is a lot.. it may make it. I don't think any insufficient top bars can crack the top in tension (like in bottom of beams in tension). Would it? Using the strut and tie analogy. Any insufficient tie in the Opening case would mereLY cause punching shear.. but if the punching shear capacity is a lot.. it may calc out.. what you think? I just need to master the concepts so I know how to debate with the designers.. so many many thanks Kootk.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
After reading many books more. I finally got your core ideas. Kootk.. I finally am confident how to present the ideas to my designers. The moments of the column has to be carried to the footing.. and moment transfer by shear stress is not enough to do it.. but has to be carried by flexure. I'll emphasize this to them tomorrow. Because they would reason 2 way punching shear can take care of both moments and axial load. I'll say this is the not the issue I'm concerned.. but the yield of the joint itself just like column flexural must be stronger than beam's flexural. Last time I tried to share them ideas of stress-strain incompatibility of concrete and pure epoxy. They won't believe it. So don't blame me if I get a bit paranoid on all this. So many many thanks to you Kootk. I think you must get a reward for great service to the public. You would make a very good educator perhaps as professor :)

RE: Horizontal Trussing Threshold for Roof Diaphragm

Ok, we'll leave it at that then mes7a. I hope that you have a fruitful discussion with your engineers tomorrow.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

Ok, we'll leave it at that then mes7a. I hope that you have a fruitful discussion with your engineers tomorrow.

Oh, last thing I wanna know before getting into formal discussions with them.

I know base shear is computed from the sum of all forces above.. however, in high rise, the flexibility of the building would produce less forces that could even make the base shear of 30-storey vs 4 storey not far. For example.

4 story building - 20m x 30 m residential in Zone -3 on a hard rock. R = 3 and I = 1.
Approx time period = 0.4 sec
Approx loading = 12 kn/m2
Total Base shear (using IS1893) = 6.7% of seismic weight = 20x30x4x12x6.7/100 = 1930 kn.

30 story building - 20m x 30 m residential in Zone -3 on a hard rock. R = 3 and I = 1.
Approx time period = 3.0 sec
Approx loading = 15 kn/m2 (as columns and walls will be much thicker)
Total Base shear (using IS1893) = 0.9% of seismic weight = 20x30x30x15x0.9/100 = 2430 kn.

As you can see, the 4 story building has relatively very high base shear (about 80% of 30-storey building).

Now I'd like to know the relationship of base moments to base shear. From base shear alone.. can you get estimated values of the base moments.. is there a direct formula that relates the two? Does building flexibility also affect the base moments.. or is the base moments related to all the beam moments above. You know yield base moments would occur from increasing base moments up to forming plastic hinges.

I'm asking all these because I wonder how big is the column-footing joint flexural demands in high-rise buildings compare to 4 storey (or 3 storey). Remember my designers handle high-rise.. so if they fail in my footing design.. they may miss it too in high-rise. And I'll get to the bottom of it (pun intended).

This is the last question for this long thread. Thanks a million Kootk!

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (mes7a)

This is the last question for this long thread.

Keep asking until you've gotten what you need. Don't worry about me; I'll withdraw if I get exhausted. I don't even consider a thread to be the real deal unless the replies get to triple digits!

Quote (mes7a)

Now I'd like to know the relationship of base moments to base shear.

First, we need some definitions:

M_e = seismic moments calculated from code procedures as you have done above. These should scale up and down in direct proportion to the base shears. Flexibility results in lower moments.

M_o/s = over strength moment of your columns that depends only on the column section propertiess (column size, rebar, f'c, etc).

In a new design, things unfold like this:

1) Find M_e
2) Design column for M_e while minimizing ratio M_o/s / M_e
3) Design footing for M_o/s.

In the above case, M_o/s will obviously rise and fall with base shear.

When analyzing an existing situation, M_o/s is whatever it is based on the column section properties and neither M_e nor the base shear will enter into the equation. This is your case I believe.


I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

Keep asking until you've gotten what you need. Don't worry about me; I'll withdraw if I get exhausted. I don't even consider a thread to be the real deal unless the replies get to triple digits!

You deserved a vacation in the Bahamas for this :)
But I'd just want to know some basics and leave it to the experts as this is not my job. My job is just to live in the building.

Quote:

First, we need some definitions:

M_e = seismic moments calculated from code procedures as you have done above. These should scale up and down in direct proportion to the base shears. Flexibility results in lower moments.

M_o/s = over strength moment of your columns that depends only on the column section propertiess (column size, rebar, f'c, etc).

In a new design, things unfold like this:

1) Find M_e
2) Design column for M_e while minimizing ratio M_o/s / M_e
3) Design footing for M_o/s.

In the above case, M_o/s will obviously rise and fall with base shear.

When analyzing an existing situation, M_o/s is whatever it is based on the column section properties and neither M_e nor the base shear will enter into the equation. This is your case I believe.

But if a structure lateral resisting sytem is composed of shear wall or braced frames.. the column bases won't reach Mpr (hence there would be no yielding).. isn't it? This is the reason why if you can alter the building main seismic resisting system, you can avoid column base plastic hinges (or would this still occur even with braced frames?)

Earlier you asked me this:

Quote:

I think that the best bet for your building is to simply abandon the special moment frame concept. Analyse the structure as an ordinary or intermediate moment frame system where the demand on the joints is much less. I believe that this would work as your building is small and has several moment frames in each direction.

But a special moment frame includes vertical lateral system that are all special moment frame and even included fixed foundation.. so why make R=5 by analyzing it as ordinary or intermediate moment frame?

Quote:

Frankly, I very much doubt that it was necessary to eliminate the upper floor of the building. Do you have any concrete elevator or stair shafts in the building?



There is no elevator shaft.. only stair shafts and it is in the lower right side of the building. It may not serve as major lateral resisting system, would it.. it may even introduce torsion. What concerns me is the transverse side main lateral resisting system of the building is purely special moment frame.. not even have walls to make it stiff or shear wall nor braced frame. For the 30-story high rise.. their lateral resisting system is stiff elevator shafts. So my building column base seemed to have much more demand than high-rise.

Again, for high rise buildings with elevator shaft as main lateral resisting system and other rigid shear wall or bracing, would the moments in the column bases in the foundation be much that can make it into yielding moments?

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (mes7a)

But if a structure lateral resisting sytem is composed of shear wall or braced frames.. the column bases won't reach Mpr (hence there would be no yielding).. isn't it? This is the reason why if you can alter the building main seismic resisting system, you can avoid column base plastic hinges (or would this still occur even with braced frames?)

This is mostly right. If you have shear walls, your column bases are unlikely to reach Mpr. If you have braced frames, the bases that are part of the braced frames will have to reach Mpr. Again, there's no other way for the braced frames to form complete hinges unless the bases yield. But yeah, this is a valid strategy for protecting your column bases from over-strength bending forces.

Quote (mes7a)

But a special moment frame includes vertical lateral system that are all special moment frame and even included fixed foundation.. so why make R=5 by analyzing it as ordinary or intermediate moment frame?

What's killing you with the special moment frames is that you're not designing for the earthquake forces at all (M_e). Instead, because you have special moment frames, you're designing for the moment capacity of your columns. If you go OMF, M_e might go up but you will no longer need to consider M_o/s.

Quote (mes7a)

only stair shafts and it is in the lower right side of the building. It may not serve as major lateral resisting system, would it.. it may even introduce torsion

With the shaft walls considered, I anticipate that the building will function as shown below where the only active moment frames are those shown in purple. The other moment frames will be prevented from participating by the presence of the relatively stiff shaft walls that prevent those frames from deforming freely. All tolled, this is probably great news for your building.

Quote (mes7a)

gain, for high rise buildings with elevator shaft as main lateral resisting system and other rigid shear wall or bracing, would the moments in the column bases in the foundation be much that can make it into yielding moments?

No. The elevator shaft walls would keep interstory drifts so low that the columns would not develop large enough moments to cause yielding. At least, that's the idea.





I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

What's killing you with the special moment frames is that you're not designing for the earthquake forces at all (M_e). Instead, because you have special moment frames, you're designing for the moment capacity of your columns. If you go OMF, M_e might go up but you will no longer need to consider M_o/s.

What? You are saying you can analyze special moment frames as ordinary moment frames? This is the most confusing sentences you have written in all :)

Are you saying that because I built only 3 storey out of the designed 4 storey and the columns are bigger than normal.. it can be considered ordinary moment frames??

Quote:

With the shaft walls considered, I anticipate that the building will function as shown below where the only active moment frames are those shown in purple. The other moment frames will be prevented from participating by the presence of the relatively stiff shaft walls that prevent those frames from deforming freely. All tolled, this is probably great news for your building.

No the shaft walls are very soft.. it is just ordinary stairs.. only the steps are rigid.. we don't build rigid stair shafts.. the walls are just thin 4" hollow blocks mildy reinforced that you can put a hole with just using hammer.. so it won't have active particular in seismic activity in the lateral sides.. it is still the columns that resist them.

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (mes7a)

You are saying you can analyze special moment frames as ordinary moment frames?

That's exactly what I'm saying. The loads go up but the requirement for over-strength design goes away.

Quote (mes7a)

Are you saying that because I built only 3 storey out of the designed 4 storey and the columns are bigger than normal.. it can be considered ordinary moment frames??

I'm saying that it's an alternative worth investigating. I would wager that it would work out even with the 4th floor.

Quote (mes7a)

No the shaft walls are very soft.. it is just ordinary stairs.. only the steps are rigid.. we don't build rigid stair shafts.. the walls are just thin 4" hollow blocks mildy reinforced that you can put a hole with just using hammer.. so it won't have active particular in seismic activity in the lateral sides.. it is still the columns that resist them.

Agreed.

Quote (mes7a)

Why.. in your country.. do you build solid shear wall stair shalf.. why would you do that??

1) So we don't need moment frames.
2) The cost of a concrete wall versus an infill wall constructed to be fire resistant seems to be about the same.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

That's exactly what I'm saying. The loads go up but the requirement for over-strength design goes away.

Hmm.. but people built special moment frames so the R would be 8.. it should behave as special moment frames.. why would it becomes ordinary moment frames in reality??

Hmm.. did you mention all this because my columns are bigger than normal (even for 4 storeys)? Because if they are small making R exactly 8 and flexible.. it won't behave as ordinary moment frames at all.

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (OP)

Hmm.. but people built special moment frames so the R would be 8.. it should behave as special moment frames.. why would it becomes ordinary moment frames in reality??

We use high R values so that we can use low design forces. Conversely, if a structure can handle higher design forces, we can use design and detailing procedures associated with lower ductility demands (low R). The only difference between an OMF and an SMF is that you can push an SMF a lot farther before it falls apart. About four times farther. Of course, if you never need to push quite so far, all the better.

Quote (OP)

Hmm.. did you mention all this because my columns are bigger than normal (even for 4 storeys)? Because if they are small making R exactly 8 and flexible.. it won't behave as ordinary moment frames at all.

I mentioned it because:

1) your columns are big.
2) your building is short.
3) you've got moment frames all over the place.
4) I know that you've lopped off a whole story.

All of these things point to reserve pre-plastic hinging capacity. Reserve strength that you can mobilize before anything is required to yield.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

I mentioned it because:

1) your columns are big.
2) your building is short.
3) you've got moment frames all over the place.
4) I know that you've lopped off a whole story.

All of these things point to reserve pre-plastic hinging capacity. Reserve strength that you can mobilize before anything is required to yield.

The reason I lopped off a while story is because of the epoxy ejected in a 4 inch thick honeycombs in one of the columns near footing. See:



If you get it to a compressive strength test. It's about 8000 PSI. The designers said it's even stronger than concrete and most designers used it here whenever there are honeycombs in the columns or beams (honeycombs are common because of extremely bad construction practice in the country). But after numerous computations with experts in eng-tips. We determined there was a loss of about 1000kN (with 2000kN left with the building 1000kN actual live, dead and sd load) capacity in the epoxy section (due to modulus of elasticity and strain incompatibility between epoxy and concrete). But the designers couldn't understand what I'm saying after showing him the computations and reminding him about stress-strain diagrams. So I gave up and just decided to go for 3 storey as safety margin.

Anyway. I talked with the designer a while ago about the flexural overstrength in the column-footing joint capacity you mentioned at length here. Again, they said punching shear is sufficient. I emphasized it's the joint flexural moment transfer that may not occur and can yield. They said the footing has been designed for punching shear. They can't seem to understand the concept of joint moment transfer of column base to footing by flexure or tie in the strut and tie model (which they don't use). But then maybe they don't really design for special moment frames which you said depends on column moment capacity. Even before designing the beams, they already had decided to make the column 0.5x0.5 (and 0.5x0.4) meter. If you will use pure special moment computations, the size of the joint and column depends on the beams probable moment strength. But it seems they don't follow it. What happens.. based on your earlier explanation is they designed it as ordinary moment frame with big columns and lower R that can resist earthquake moments and base shear. This is the only logical thing possible based on what they described and what you described. Whatever, even with ordinary moment frame anywhere, if you have magnitude 9 earthquakes. The column-footing joint would go into overstrength flexure (forming plastic hinges) isn't it? Or would there be shear failure first before plastic hinges forming in column base in ordinary moment frames at maximum earthquake magnitude (say 9)?

Anyway. I just plan to building up to 3 storey with light roof (and possibly light wall). But now I'm worried the resonance of the soil/rock underneath can match the period of the building. When there is resonance.. how many times would the seismic movement be compared to non-resonance case? Because if the 3-storey with light roofing would be in resonance with the soil. It would be worse than having heavier roofing. And should there be resonance.. I'd use heavier roofing like steel deck and heavier wall to desynchronize the resonance. Any experience with computing for soil and building resonance effect? Don't blame me for being paranoid. I don't know where to find local designers who know about the epoxy strain incompatibility and the column base-footing joint flexure capacity that international experts in eng-tips know about. So I'm very cautious now of everything. Again Thanks a zillion.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
To continue what I said in middle paragraph above.

Quote:

We use high R values so that we can use low design forces. Conversely, if a structure can handle higher design forces, we can use design and detailing procedures associated with lower ductility demands (low R). The only difference between an OMF and an SMF is that you can push an SMF a lot farther before it falls apart. About four times farther. Of course, if you never need to push quite so far, all the better.

I've been thinking of this for hours. And this is the confusing part. If the structure can handle higher design forces.. yet you use detailing procedures associated with lower ductility demands (low R), then would there be ductility or not.. We design special moment frames to resist higher seismic load because the yield displacement can absorb a lot of energy.. hence the "ductility" part.. yet you said above about using detailing procedures with lower ductility demands in structure that can handle design forces.. which seems to be in contradiction because when you use low ductility detailing, shear failure can occur from beam plastic hinges. I'm asking this because my beams were heavily retrofitted with carbon fiber and so many hoops in the hinging ends. They are designing for higher ductility. Would they be mobilized when the elastic limit of the bigger columns have been used up? If yes.. then the right words to say is "Conversely, if a structure can handle higher design forces, we can use design and detailing procedures associated with HIGHER ductility demands (High R) too." Then you will end up with an Adaptive R and Hybrid OMF/SMF where they can be push 4 times farther also. Is it not?

Quote:


I mentioned it because:

1) your columns are big.
2) your building is short.
3) you've got moment frames all over the place.
4) I know that you've lopped off a whole story.

All of these things point to reserve pre-plastic hinging capacity. Reserve strength that you can mobilize before anything is required to yield.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

By the way. We are less than 10 kilometers away from a major fault line that can create magnitude 7 earthquake in the capital of our country. Therefore I think it's not a good idea to consider it going back to Ordinary Moment Frame. I think the better description would be special moment frames all along. So what if my columns are bigger. It's just like the case of high-rise where the lower storey has bigger column.. it can only attract the moments away from beams. But it's all special moment frames all the way. Do you agree with this? Let's call it Adaptive R then with columns that can work initially from lower R but then go to higher R as seismic activity increases in magnitude enough to make the columns drift into ductile mode (and consequent plastic rotations of the beams). ??

RE: Horizontal Trussing Threshold for Roof Diaphragm

If a building had no ductility, then we would design it for an R value of 1. There is inherent damping in the system, which is why we use R values to reduce the seismic load. In high R value systems, we use special detailing to allow plastic hinges to form in certain locations, which further dissipates energy and leads to a higher R value. In low R value systems, like ordinary moment frames, we do not pay special attention to detailing because it is often times very expensive and the loads are not large enough to require the special detailing. In fact, steel special or intermediate moment frames have compactness criteria that limits the size of beams and columns that can be used. So in a low seismic region, it is not economical to use a higher R value system. It's important to note that most design codes will not allow the use of low R value systems in regions with high seismic potential. Also, the R value doesn't change as seismic activity increases. You design and detail for one force, usually what is expected from a 500 year mean recurrence period for seismic events.

All that being said, I think you have a good basic understanding of what the goal is in seismic design. However, I think you should review some basic concepts within structural dynamics and how seismic loads travel through the building to get a better understanding of designing for high seismic regions. These are basic concepts of structural engineering for lateral design and are critical to understand in high seismic regions. Most of your questions can only be answered by modeling and applying loads to see how the frames behave. It's not very intuitive (for most) to predict what happens in a frame as you started changing detailing and frame proportions.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
I've been reading Iranian literature on it to getting a handle of what Kootk is saying.

http://www.academia.edu/10000462/The_Economical_Ef...

It concludes: Selecting a higher ductility level increases transverse reinforcements and then it will increase the weight of these bars. However, on the other hand, for ductile frames we can consider small seismic loads, which consequently decrease the longitudinal reinforcements.

Do ductile frames (special moment frames) have lower longitudinal reinforcements than ordinary frames?

Kootk golden words were "We use high R values so that we can use low design forces. Conversely, if a structure can handle higher design forces, we can use design and detailing procedures associated with lower ductility demands (low R). The only difference between an OMF and an SMF is that you can push an SMF a lot farther before it falls apart. About four times farther. Of course, if you never need to push quite so far, all the better."

For ordinary moment frames that deal directly with the seismic forces (M_e). Can't it be make ductile too? Or does increase longitudinal bars used in ordinary moment frames means they can't be made ductile?? Is this what Kootk mean?

I'm afraid my building which has been designed to resist maximum earthquakes with less than 10km from a fault line has been turned into ordinary moment frame instead of special moment frames unfortunately. Is this it? Maybe making the columns bigger create this unfortunately scenario of weakening seismic resistance because it has to face large seismic forces (bec not flexible.. making R less)

RE: Horizontal Trussing Threshold for Roof Diaphragm

The size and number of longitudinal bars are a function of the force going into them. It all depends on frame configuration and the forces (gravity and lateral) at each frame.

If you do not meet the detailing requirements for a special moment frame, then you do no have a special moment frame.

RE: Horizontal Trussing Threshold for Roof Diaphragm

I made you a spiffy sketch mes7a (below). If you study it carefully, it will reveal a fundamental principle that is built into modern seismic design provisions but isn't discussed much. That principle ss called the Equal Displacement Principle. If you google that term, you'll turn up a ton of related explanatory material.

The gist of the equal displacement principle is that a building lateral system will experience approximately the same amount of displacement under design seismic excitation regardless of the level of ductility built into the system. Whether it's R=2 or R=8, the total displacement is the same. The only difference is that an R=8 system would experience more of that displacement as plastic deformation and less as elastic deformation.

In the end, an R=2 building will dissipate just as much seismic energy as an R=8 building. It will just take more oscillations to do it. In the example shown below, the R=2 building wold require 2.29 times as many oscillations to dissipate the same amount of energy as the R=8 building.

The take away from all this is that, fundamentally, a low R building is no more dangerous seismically than a high R building so long as the design seismic forces are adjusted appropriately. Of course, if your local building code flat out prohibits low ductility systems, then that is another matter. The one difficulty with low R design is that it's very difficult to predict seismic loads accurately. And high R systems tend to be more forgiving. That's why some codes prohibit low R systems in high seismic regions.

Sometimes essential services buildings in high seismic regions will be intentionally constructed using low R systems. That is done because a properly designed low R system can be expected to suffer the least amount of damage during an earthquake. R=8 means that a building doesn't fall down but it will be nearly destroyed after a design seismic event. Such a building may have compromised vertical transportation and mechanical systems and may need to be demolished rather than re-occupied. In theory, an R=1 hospital should never even have to close its doors.

In New Zealand, they're beginning to question whether or high ductility is even a good idea. Poor Christchurch was damaged so badly that they even considered relocating the city rather than attempt to repair all of the damaged infrastructure. The modern trend towards performance based design is taking us in this direction as well. Owners of buildings where the contents are far more valuable than the building (data centres etc) don't want their facilities to have to weather R=8 damage.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

The take away from all this is that, fundamentally, a low R building is no more dangerous seismically than a high R building so long as the design seismic forces are adjusted appropriately. Of course, if your local building code flat out prohibits low ductility systems, then that is another matter. The one difficulty with low R design is that it's very difficult to predict seismic loads accurately. And high R systems tend to be more forgiving. That's why some codes prohibit low R systems in high seismic regions.

Our codes copy everything 100% from ACI. But building officials engineers don't focus on differentiations between low and high R so they just gave approval for gravity load. Therefore designers in my place are exempted from any liability should the building suffer from any seismic activity.

Anyway.



For Low R building with detailing that matches special moment frames.. like my structure which matches the detailing of special moment frames. Then it's still called Low R because the column are more rigid.. isn't it. Failure mode of ordinary moment frames are shear failure. So with my shear detailing of columns at 70mm o.c. at upper and lower part, then the effect is more capacity elastically? Now the question is.. you said you can push special moment frames 4 times more than ordinary moment frames. But what happens if the ordinary moment frames has shear detailing just like SMF and even more flexural capacity than SMF (the paper above says SMF has fewer flexural rebars so it can yield taking part of the seismic force).. then can the OMF be pushed 4 times too and perform like SMF?

Also when you are building a say 4 storey building with designed R=8.. and you just do actual 2 storey.. is it like turning the R into 5? In details, the effect of is like if you build the full R-8 compliant 4 storey. The columns became flexible enough (R=8).. but if you just finish 2 storey, then the columns became rigid in comparison with other 2-storey (then R=5). Is this concept right? This means if I finish the designed 4 storey. Then it becomes bonafide special moment frames? If I just finished 2 out of 4 storey. It's become ordinary moment frames (as far as flexibility is concerned) or no difference at all?

RE: Horizontal Trussing Threshold for Roof Diaphragm

You're heading off the rails here mes7a. Think of it like this:

1) If all parts of your building were SMF compliant, we wouldn't be having this conversation.

2) Some parts of your building are SMF compliant (columns) and others are not (footing connections).

3) Because not all parts of your building are SMF compliant, you do not have an SMF (mike's point).

4) What you're left with could be conceived of as an OMF or IMF in which some members, but not all, posses more ductility than they require in order to satisfy OMF/IFM requirements. Good for them.

5) In no case would SMF detailing cause an otherwise acceptable member to be unsuitable for use in an OMF/IMF.

Quote (mes7a)

then can the OMF be pushed 4 times too and perform like SMF?

No. Once drift reached the OMF limit, the portions of your structure that are not SMF compliant would fail. Your frame is, essentially, only as ductile as the least ductile component.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

You're heading off the rails here mes7a. Think of it like this:

1) If all parts of your building were SMF compliant, we wouldn't be having this conversation.

2) Some parts of your building are SMF compliant (columns) and others are not (footing connections).

3) Because not all parts of your building are SMF compliant, you do not have an SMF (mike's point).

4) What you're left with could be conceived of as an OMF or IMF in which some members, but not all, posses more ductility than they require in order to satisfy OMF/IFM requirements. Good for them.

5) In no case would SMF detailing cause an otherwise acceptable member to be unsuitable for use in an OMF/IMF.

Ok. Since there is insufficiency (like the footing connection). Then if I jus finished 2 storey out of 4. The OMG part would be stronger and has more elasticity than building the full 4 storey right?

Because if the effect is significant. I just just retain the 2 storey and treat it as very strong OMF with more elastic limit than a full 3 or 4 storey OMF that is weaker elastically.

But generally it's really true that if you build just half of a special moment frame building storeys.. the columns would be more rigid for the design and become an OMF?

RE: Horizontal Trussing Threshold for Roof Diaphragm

Think of it like this:

1) any given level of your building is capable of a certain, safe, total lateral displacement of which part is elastic and part is inelastic. Lopping off stories doesn't change this.

2) lopping off stories will result in your building experiencing less lateral seismic displacement. This is the only thing that lopping off stories will affect.

3) with less demand for seismic lateral displacement, there is a greater likelihood that the displacement demand (#2) will not exceed total displacement capacity (#1). This is the most fundamental way of stating the condition of seismic safety.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

Think of it like this:

1) any given level of your building is capable of a certain, safe, total lateral displacement of which part is elastic and part is inelastic. Lopping off stories doesn't change this.

What I'm saying is simply this. Imagine you are building a special moment frame 30-storey building with 2 meters diameter columns at ground floor.. but you ran out of budget and only build 2-storey out of it. The 2 meters wide columns sure won't be called special moment frame, isn't it? because it won't be flexible. So in this case.. do you called the 2 storey building (out of 30-storey design) ordinary moment frame or special moment frame?.

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (mes7a)

The 2 meters wide columns sure won't be called special moment frame, isn't it? because it won't be flexible.

No, the columns would still be SMF compliant. SMF is about ductility, not flexibility. Column ductility is a function of the column properties and nothing else.

Quote (mes7a)

. do you called the 2 storey building (out of 30-storey design) ordinary moment frame or special moment frame?.

I would call it a grossly oversized SMF unlikely to ever see lateral displacement ductility demands exceeding those associated with OMF.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
Thanks Kootk. This would explain my paranoia.

http://www.tempo.com.ph/2013/10/17/7-2-earthquake-...

"With a magnitude of 7.2, the study found that 170,000 residential houses will collapse, 340,000 residential houses will be partly damaged, 34,000 persons will die, and 114,000 persons will be injured."

My building is less than 10 kilometers away from the fault line. Our capital would get back to the stone age should the west valley quake occur.

So my paranoia is really vigilance.

In your building design... do you compute for the soil/rock filtering of accelerations or waves? because if it matches the building fundamental period.. the resonance can increase the seismic force.. as estimate.. how many times do you think is the effect should there be resonance of soil/rock acceleration and building period?

RE: Horizontal Trussing Threshold for Roof Diaphragm

I would have to rely on geotechnical engineers for guidance with the soil resonance. I'd recommend posting this question in that section of the forum. It's not considered explicitly in routine design. To some extent, it's built into the seismic site classification system I believe.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

Holy smokes!

"It is imperative Cunth doesn't get his hands on those codes."

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

No, the columns would still be SMF compliant. SMF is about ductility, not flexibility. Column ductility is a function of the column properties and nothing else.

Professor Kootk. Thanks for this very important distinction or clarifications. Because in the article "Seismic Design of Reinforced Concrete Special Moments Frames: Guide for Practicing Engineers". it is stated that:

"Both strength and stiffness need to be considered in the design of special moment frames. According to ASCE 7, special moment frames are allowed to be designed for a force reduction factor of R = 8. That is, they are allowed to be designed for a base shear equal to one-eighth of the value obtained from an elastic response analysis. Moment frames are generally flexible lateral systems; therefore, strength requirements may be controlled by the minimum base shear equations of the code."

This gave me the impression for a couple of years that special moment frames is about flexibility.

In Wight book. It is stated:

"As discussed previously, more ductile structural systems can be safetly designed for lower seismic forces than systems with limited ductility." The Equal Displacement Principle is shared there as:



But then they say not to make the lateral system so stiff so as not to attract more seismic forces. Isn't this the same as flexibility? Or can you really make the columns much bigger and don't care about the seismic force it can attract because shear can easily be handled by columns via lateral ties.. this means as long as the columns have enough ductility.. then it's ok to make it bigger.. and have at least more elastic reserve before ductile response initiated and this won't take away the SMF label. Is this what you were pointing out?


RE: Horizontal Trussing Threshold for Roof Diaphragm

Stiffer systems often do attract more seismic load. That is indeed true. However, ductility and stiffness do not always vary in unison. Just because something is stiff, that doesn't mean that it isn't also ductile. You can have both very stiff SMF and very flexible SMF. And the same is true of OMF.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

Stiffer systems often do attract more seismic load. That is indeed true. However, ductility and stiffness do not always vary in unison. Just because something is stiff, that doesn't mean that it isn't also ductile. You can have both very stiff SMF and very flexible SMF. And the same is true of OMF.

Ok. And while talking about the article read by thousands: "Seismic Design of Reinforced Concrete Special Moments Frames: Guide for Practicing Engineers". It is mentioned in the foundation part that:

"Modeling pinned restraints at the base of the columns, Figure 4-1 (a), is typical for frames that do not extend through floors below grade. This assumption results in the most flexible column base restraint. The high flexibility will lengthen the period of the building, resulting in a lower calculated base shear but larger calculated drifts. Pinned restraints at the column bases will also simplify the design of the footing. Where pinned restraints have been modeled, dowels connecting the column base to the foundation need to be capable of transferring the shear and axial forces to the foundation."

But you mentioned special moment frames footing must be fixed to create complete mechanism. Why would the article talked about pinned restrains which won't create any complete mechanism (plastic hinges at the column bases in the footing)?

Maybe because when your column base form plastic hinges.. the building is no longer repairable. So some engineers can get away with pinned footing restrained? What do you make of it? Why would the article mentioned about pinned footing if it is not part of Special Moment Frame which is their main topic?

RE: Horizontal Trussing Threshold for Roof Diaphragm

You only need column base plastic hinges to form a mechanism if you have modelled your column bases as fixed. The whole idea is to create a true pin at the column base. If you started off with a pinned column base, this step is already complete.

I disagree with the referenced paper regarding whether or not column bases at grade level pad footings should be modelled as pinned. My opinion is that typical construction details at column/firing joints produce considerable fixity that should be accounted for. That fixity must be overcome before mechanism formation is complete.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

The take away from all this is that, fundamentally, a low R building is no more dangerous seismically than a high R building so long as the design seismic forces are adjusted appropriately. Of course, if your local building code flat out prohibits low ductility systems, then that is another matter. The one difficulty with low R design is that it's very difficult to predict seismic loads accurately. And high R systems tend to be more forgiving. That's why some codes prohibit low R systems in high seismic regions.

You mentioned above "some codes prohibit low R systems in high seismic regions". All I heard of or read is ordinary moment frames can't be used in high seismic regions (our country copied the ACI codes word by word). What particular codes don't forbid them? I want to read about them.

I think an R=1 is something that can respond elastically like nuclear power plants. So these vital installations use R=1 and ordinary moment frames, huh?

RE: Horizontal Trussing Threshold for Roof Diaphragm

I don't have answers to either of those questions mes7a. Perhaps other forum members will be able to comment.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

I don't know what codes don't forbid OMF in high seismic.

I don't think using R=1 necessarily means an elastic design. There is a method to keep critical structures close to elastic in an EQ, but it involves a time-history suite and R=0.85 if I remember correctly. I will try to dig up the reference.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

Stiffer systems often do attract more seismic load. That is indeed true. However, ductility and stiffness do not always vary in unison. Just because something is stiff, that doesn't mean that it isn't also ductile. You can have both very stiff SMF and very flexible SMF. And the same is true of OMF.

"Stiff SMF" as you mentioned above may no longer have R=8. There must be terms to refer to Stiff SMF that have R=5. Or maybe Stiff SMF must be called IMF that is ductile enough to have adaptive R from 5 to 8?

Also I read this:

https://www.ideals.illinois.edu/bitstream/handle/2...

"They concluded that for a structure of a natural period less than 0.2 second (short period structures), the ductility does not help in reducing the response of the structure. Hence, for such structures, no ductility reduction factor should be used."

Structure having natural period of 0.2 is about 2 to 3 storey building. So it's possible such buildings are so stiff that strength must be priorized? Because if you make 2 storey building so flexible.. it may be in danger of P-delta effect or sorta.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
Dear Kootk..

Let me share something which have been bothering me for years and related to my massive inquiries about moment frames and lessening drift and beam rotations.. it's because my beams were not really special moment frames. Here's why.

In our country. Our designers use stirrup spacing not from calculations but from memorization of 10mm at 4 inch o.c. and larger spacing as it goes to midspan.. or from uniform load.

When the designers design my girder with secondary beam framing into it. They use stirrups spacing of uniform load. Later I told him after construction is finished that his spacing is wrong (as commented by others). Then he realized it's true. The shear should not be from uniform load but instead look like this:



Anyway. For secondary beam framing into primary. Does the shear cracks look like the following?



Indian structural engineers I consulted said the cracks occur 2D away from the point where secondary frame into primary.
Is it true? My fear is it may be closer to the center.

This is another reason I didn't continue the top floor.. to lessen story shear below.. and hoping the bigger columns can prevent drift and beam rotations. Their computations show the midspan is good for gravity loads. And they recommended carbon fiber for the missing capacity for cyclic loading. The designers are so convinced the carbon fiber can make up for insufficient stirrups at midspan but I read carbon fiber doesn't work much.

So my beams and footing connection are not special moment frames. My hope is the bigger columns and lesser storey can create the scenario where the columns can become OMF or IMF with ductility and drifts would be less enough to avoid cyclic loading in the beams.

Again my main question is. During cyclic loading.. would the cracks at midspan where secondary beam frames into it be 2D away from midspan or nearer? What is your experience of this? I got nervous for it for months about it a year back and still now.

RE: Horizontal Trussing Threshold for Roof Diaphragm

Your concern is whether or not the shear cracks will be crossed by the FRP stirrups, right? If so then, yes, your cracks should emanate from the top of the beam and emanate downwards to cross the FRP stirrups.

The shear detailing of your beam would satisfy ACI requirements. In my jurisdiction, however, the practice is to provide "hanger" reinforcement stirrups right behind the supported beam. See the sketch below. This practice has always been debatable, however, as zillions of beams get cast in the US without hanger (suspension) reinforcement and nothing terrible ever seems to happen as a result.





I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

Your concern is whether or not the shear cracks will be crossed by the FRP stirrups, right? If so then, yes, your cracks should emanate from the top of the beam and emanate downwards to cross the FRP stirrups.

But FRP doesn't work. I have spent over 2 months reading about it 2 yrs ago. It can debond from surface before it breaks.. so I don't trust it. Instead I trust on the original stirrups. If the cracks angle is like you described.. some of the original rebar stirrups can cross it. But the problem is that in seismic movement and cyclic loading.. the plastic hinge can form at midspan and the shear cracks can be anywhere. See:




This is for beam with uniform span in cyclic loading. Do you have picture of how it looks like when a secondary beam frames into the primary and it is in seismic cyclic loading? Where would the plastic hinge form? Right at midspan? Then the lack of original rebar stirrups worry me at times.

This is why I'm hoping not building the 4th storey would lessen storey shear below and make the building stiffer and just save it.

Anyway. Don't worry. We would stop at the 100th message. Lol. I know it's a long thread. And at 100th message. I'd stop and reflect about it all.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

In my jurisdiction, however, the practice is to provide "hanger" reinforcement stirrups right behind the supported beam. See the sketch below. This practice has always been debatable, however, as zillions of beams get cast in the US without hanger (suspension) reinforcement and nothing terrible ever seems to happen as a result.

About hanger reinforcement, I actually got worried about it 2 years ago and discussed it with indian structural engineers. The figure is from "Leonhardt, F. and E. Mönnig 1977, Vorlesungen uber Massivebau, Dritter Teil, Grundlagen zum Bewehren im Stahlbetonbau, Dritte Auflage, Springer-Verlag, Berlin, p. 246"



It seems when your secondary beam shear is more than Vc and the it can suffer diagonal crack that you really need to worry much. The drawing of the compression fan (from Mcgregor presentation of the 1984 Canadian Concrete Code 6-29)



It seems internal Hanger stirrups are only required if V (shear) at the end of the supported beam is beyond a certain threshold that can cause diagonal cracking. Dolan Et Al mentioned "Hangers will also be unnecessary if the factored beam shear is less than 0.85Vc (as is usually the case for one-way joists, for example), because in such a case diagonal cracks would not form in the supported member. The predictions of the trust model would thus not be valid, and the reaction would be more nearly uniform through the depth."

Another author Mcgregor stated " These provisions can be waived if the shear, Vu2, at the end of the supporeted beam is less than 3*sqrt(fc')bd, because inclined cracking is not fully developed at this shear."

But then it seems you are talking about noncrack secondary beam with the primary needing to transfer the reaction up.

The stirrups can equilibrate the top and bottom bars near midspan.. so it's my hope 2 years ago the beam won't just fall down. But then during reverse seismic loading (up and down). It may stress it. One more reason I want to lessen seismic load of building.

Thanks for reminding it. Because I have about 1-2 inches of concrete topping in the roofdeck now above my 2nd floor to make rain flow to drains.. But since the third floor will be covered by light wall and light roof soon. I think I'll remove all the 1-2" of concrete topping above waterproof to lessen the SD load and compensate for missing hanger reinforcement.

Anyway. In case I'll build another building. I'll surely have massive internal hanger stirrups to complete the truss analogy (stirrups only cost little so most plans must include it).

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
Kootk.. I had difficulty sleeping last night because something kept bothering me. I have thought of it 2 years ago but it's coming back to me.

In the picture you showed above:



The secondary bottom bars are resting above the primary bottom bars. There are 4 bottom bars which matches my girder's bars too.

My worry is could the bottom bars just fracture from the weight of the secondary beams (since there is lack of hanger reinforcement you mentioned.. although I had one internal hanger inside (10mm hoop). How do you exactly compute for the bar sideway tensile strength.

Using area of 20mm rebar as 0.000000314 and tensile strength of 414 MPA.. if you multiply the two, you come up with 130 kN. If there are 4 bars.. total capacity is 520kN? But it is sideway. How would you convert it to sideway capacity of a rebar?

Do you think all the load are being concentrated on the 4 bottom bars of the primary girder beam? If not.. how many percentage roughly? How do you analyze it?

My equivalent axial load of 2 secondary beam framing into the primary girder is 350kN.. Do you divide this by two to distribute it to the 2 bottom bars of the secondary beams framing into the primary?

Let's wrap it up when we reached message #100. Lol..

RE: Horizontal Trussing Threshold for Roof Diaphragm

ACI Vc shear procedures encompass a variety of things that contribute to shear resistance including friction at the compression block, aggregate interlock, and longitudinal bar dowel action. While the dowels add something to the shear capacity for certain, I wouldn't want to rely on them as the sole means of shear transfer.

Quote (mes7a)

Using area of 20mm rebar as 0.000000314 and tensile strength of 414 MPA.. if you multiply the two, you come up with 130 kN. If there are 4 bars.. total capacity is 520kN? But it is sideway. How would you convert it to sideway capacity of a rebar?

You've used dowel tensile strength here when the relevant parameter would be dowel shear strength. In reality, it isn't the rebar that fails in dowelled connections. Rather, it is the concrete surrounding the rebar. In this case, the girder concrete below the dowels might spall off. There would be dowel bending at play as well but there's probably no need to get into that.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

You've used dowel tensile strength here when the relevant parameter would be dowel shear strength. In reality, it isn't the rebar that fails in dowelled connections. Rather, it is the concrete surrounding the rebar. In this case, the girder concrete below the dowels might spall off. There would be dowel bending at play as well but there's probably no need to get into that.

Why do you call it dowelled connection. I'm talking about the 4 main longitudinal bars of the primary girder beam. The reactions from the secondary beam is towards the bottom as it framed into the primary beam. This is why you mentioned about hanger reinforcement.. to carry the force to the upper longitudinal bars of the primary girder.

Now i'm not talking about hanger reinforcement to transfer the force from bottom of girder to top.. but the shear strength of the longitudinal bar itself (it can fracture when it can no longer take the axial load, isn't it?). What is the percentage of the shear strength of the bar compared to its tensile strength? I heard it is 3/4 or 1/2? You mean the reactions from the secondary beams don't focus on the girder bottom rebars but still carried in the whole section of the concrete? To what extend.. how many percentage approx. does the rebar carry the whole axial load of the secondary beams?

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (mes7a)

Why do you call it dowelled connection.

What you've described, in my mind, is a dowelled shear connection.

Quote (mes7a)

What is the percentage of the shear strength of the bar compared to its tensile strength? I heard it is 3/4 or 1/2?

0.75 Fy at ultimate; 0.58 Fy at yield.

Quote (mes7a)

To what extend.. how many percentage approx. does the rebar carry the whole axial load of the secondary beams?

I don't know the ratios I'm afraid.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

0.75 Fy at ultimate; 0.58 Fy at yield.

I don't know the ratios I'm afraid.

About 2 years ago I was discussing it with some indian structural engineers in this indian thread:

http://www.sefindia.org/forum/viewtopic.php?t=1436...

Someone there said "The secondary beam reaction will be cusing a crack on main bar if the main bar can not resist it and if it has not been provided with extra stirrups on either side at a distance "d" and in addition as one of the sefians posted crank or bent up bar is provided."

Kootk.. do you believe it is possible the secondary beam reaction can cause a crack in the main bottom bar? Do you compute it at 0.75 Fy?

But the joint being monolithic.. hopefully the concrete can carry some load.. would it.. how do you analyze this in your design?

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (Someone)

The secondary beam reaction will be cusing a crack on main bar if the main bar can not resist it and if it has not been provided with extra stirrups on either side at a distance "d" and in addition as one of the sefians posted crank or bent up bar is provided.

I believe that the referenced cracks are cracks in the concrete surrounding the rebar rather than cracks in the rebar itself. Rebar yielding should precede rebar rupture and concrete rupture should precede rebar yielding.

Quote (mes7a)

hopefully the concrete can carry some load.. would it.. how do you analyze this in your design?

It would. I analyzed it using the hanger steel provisions or, in unusual situations, with strut and tie.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

I believe that the referenced cracks are cracks in the concrete surrounding the rebar rather than cracks in the rebar itself. Rebar yielding should precede rebar rupture and concrete rupture should precede rebar yielding.

Yes. This is what I need to know. Last night I couldn't sleep because pondering if rebar can also suffer brittle shear failure or would it yield first then break. But then if you hammer a rebar on the side. It surely can break before yield.. no?

So the Indians may be saying that without the stirrups close to the interface.. the bars may bend and then yield and then break. Judging by his tone I thought he meant it would just break brittle.

But then for higher yield rebar. It is more brittle.. it doesn't go into hardening and strain is less. In this case, the bar can literally break.

By the way. Before we wrap up the thread.. there is something the indian said but I didn't follow up 2 years ago and don't understand. He said:

"With beams framing plan as per sketch, irrespective assumptions made for design, this system is really a grid beams with ends of some beams rest on columns[ unyielding support] and some on beams [ yielding supports, since at the point deflections will occur]. This will lead to additional shear in 6m span beams resting on beams. "

What is other terms for "unyielding support" and "yielding supports"? Is he talking about rebar yielding or support conditions? He refers to yielding supports connected to deflections.. why would it be have additional shear in the secondary beams resting on primary beams? What is the equivalent concept in ACI? They may use other terms in their indian codes.

Many many thanks.

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (mes7a)

But then if you hammer a rebar on the side. It surely can break before yield.. no?

It shouldn't unless it's extremely cold, overly cold worked, embrittled via improper welding, containing manufacturing defects etc...

Quote (mes7a)

"unyielding support" and "yielding supports"? Is he talking about rebar yielding or support conditions?

Support conditions. "Stiff supports" and "felxibile supports" would have been more apt terms.

Quote (mes7a)

He refers to yielding supports connected to deflections.. why would it be have additional shear in the secondary beams resting on primary beams?

I'm afraid that I don't understand the referenced statement. Monolithic concrete structures are very indeterminate. As such, it's always tough to know how much load is going to the various elements.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

I'm afraid that I don't understand the referenced statement. Monolithic concrete structures are very indeterminate. As such, it's always tough to know how much load is going to the various elements.

His complete sentences is "This will lead to additional shear in 6m span beams resting on beams. Since length of both beams together is 12m, bottom bars may have been lapped at the junction. This lap shall be not less than Ld/3 or depth of beam, whichever is larger. This may be one of the reasons.
Cracks may occur anywhere where the deficiency is."

He doesn't know I use 12 meters long bars so there is no lap. I think he meant there may be more shear if the lap is not Ld/3 at critical portions owing to insufficient development length.

Anyway. I think that's it. I'll pause, ponder and reflect every word you said in this long thread and others. You patience is top notch. I don't know how to thank you. But thanks a million anyway for the many responses in this thread. You are incredible Kootk. Someday if you can reveal to us your real name, better so we can nominate you as educational system official! Lol.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

You can check the capacity using strut and tie procedures applied to the opening and closing models that I sketched above. Another way to look at it is to flip the model upside down in which case it becomes much like a roof level slab to column connection. The 270 kNm that I mentioned earlier was a ballpark number based on having to yield 8-20M column bars to over strength at a lever arm of 300 mm. I'd come up with a detailed estimate of your own before investing too much time on it.

Btw.. Kootk.. something important to ask...

What do you mean by "lever arm of 300mm"? Did you measure it from rebar to rebar at sides? Again, the columns at the sides are 0.5x0.4 mtr.. not 0.5x0.5 mtr (which is only used at center).



Please just focus on the C1 which is used on all the sides and corner except one.

Where did you come up with 300mm?

I have spent the year 2013 in manual computations of columns and beams to verify loadings. In the following:



You can see only half of the bars are used in moments in tension side (the compression side bars are to add to the compression block).

So when the column footing connection developed plastic hinges.. it is only one side of the column that would yield.. so since there are only 6 bars that would yield.. then one has to calculate the contribution it alone right? (because you mentioned above that "having to yield 8-20M column bars to over strength at a lever arm of 300 mm".. why do you have to use all 8-20M? or all 6-20M since we are talking of the 0.5x0.4 mtr C1)?

Also would it even get to yield before the compression block on the opposite side of the tension crushes?

My designers still told me the punching shear alone is enough without having to think about flexural capacity of the foundation.. so I'll plan to manually compute for the polar moment of inertia and see if the flexural capacity would be enough. Don't worry. I won't use it on new structure but just to verify my own existing building. Thanks so much again.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
Haynewp

Quote:

I don't know what codes don't forbid OMF in high seismic.

I don't think using R=1 necessarily means an elastic design. There is a method to keep critical structures close to elastic in an EQ, but it involves a time-history suite and R=0.85 if I remember correctly. I will try to dig up the reference.

I read R=1 means there is no ductility or sorta.. I tried to find the reference about this method you mentioned to keep critical structures close to elastic in an EQ that involves a time-history suite and R=0.85 but couldn't find it. If you dug it up.. please share it (ok in other thread as this is quite long already and I know need to be closed).

I just want to know how big are the moment demands in columns for them to perform elastically and how big they should be.. and what kind of analysis it is. Thank you.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

You can check the capacity using strut and tie procedures applied to the opening and closing models that I sketched above. Another way to look at it is to flip the model upside down in which case it becomes much like a roof level slab to column connection. The 270 kNm that I mentioned earlier was a ballpark number based on having to yield 8-20M column bars to over strength at a lever arm of 300 mm. I'd come up with a detailed estimate of your own before investing too much time on it.

Oh Kootk. I got confused for a while as I let every word you said sink in. I thought your ballpark number estimate was based on 20-20M (which I erroneously read)... but it's 8-20M.. so you are still right (I was hoping you were wrong somewhere and the footing foundation flexural transfer can be done. Although your lever arm of "300" may really be 400. I'd review the computations in our favorite textbook next week.

Whatever. If the axial load is below the interaction diagram balanced point. The column base would yield even when moments is not so much.. this means the footing-foundation connection won't be engaged as the column would fail already first (yielding to breakage). And when it is above balanced point.. the concrete would crush first... therefore consider the column flexural as possible limiting factor that can already fail before it can engage the flexural overstrength yielding lines of the footing column interface. Here there are 2 parts that are yielding independently (column tension side and the connection flexural side). In the future if found a good reference about this. Please share it.. thanks a whole lot.

RE: Horizontal Trussing Threshold for Roof Diaphragm

Yeah, my ballpark estimate of the column flexural over strength was just that: a ballpark estimate. I used my judgement to guess at the effective lever arm and the number of tension bars participating. For real checks, you'd want a better, independently calculated estimate.

For a special moment frame, fixed based columns have to be able to yield at their bases without brittle shear/crushing failure. As such, the footing would never be prevented from experiencing the column over strength moments.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

Yeah, my ballpark estimate of the column flexural over strength was just that: a ballpark estimate. I used my judgement to guess at the effective lever arm and the number of tension bars participating. For real checks, you'd want a better, independently calculated estimate.

For a special moment frame, fixed based columns have to be able to yield at their bases without brittle shear/crushing failure. As such, the footing would never be prevented from experiencing the column over strength moments.

If the axial loads of my column bases at foundation is very low or a bit far from the balanced point. Low moment or low seismic activity would already yield it. This means the footing would take lower flexural strength to experience the column overstrength moments. Right.

But then when my column bases yielded at low magnitude 6 earthquakes and the foundation remain intact.. because it experiences lower moments.. remember the reason for it experiencing much less moment is because when the column tension side yielded, the compression block is not at maximum compression (less moments) and remembering the compression side block of a column section also participates in moments. Therefore I think a good strategy is to determining where is the axial load of my existing structure to see how far or near or if it is in the balanced point isn't it.. so moment capacity would increase.. therefore if my present 2 storey with roofdeck is much lower than the balanced point.. adding light roof won't change it much.. then have to finish the original 4 storey concrete building.. perhaps the axial loads would be in the balanced point or near it.. is this a good way to think? At balanced point moments with higher moment capacity. if the Big One (Earthquake once in 200 years) really came.. and the footing separates from lack of complete moment flexural overstrength.. well it's as bad as the building experiences complete mechanism.. noting the soil/rock underneath is much harder than designed (existing foundation is 3 times bigger.. the designers admit.. so in the even the footing separates.. the rock below has enough bearing pressure to support the separated column base... the point is.. in both situations.. they would both be useless (separated column bases or complete hinging mechanism formed)anyway after a major seismic activity.. so need to demolish and building new one. What you make of all this especially making sure the building axial loads need to be checked and be made close to the balanced point of the interaction diagram for more moment capacity and seismic resistance until footing separate or complete hinging mechanism?

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
to continue.. unless you are saying that even if column moment is much lesser than the balanced point, if the seismic demands push the moments to the balanced point.. the footing would still experience that moment (of the column base as a whole) even if the tension part of the column is already yielding? In other words, the column base foundation flexural overstrength capacity is depending on the column moment as a whole and not on the tension part yielding moment of the column.. isn't it. Point is if the tension part yields, the column can still bend up to the balanced point? This is your point isn't it.

So irregardless if the axial load is really in the balanced point or much below.. the column moment would still reach the balanced point and this is where your column base footing flexural overstrength comes in? I was thinking in last message that if the moment is much lower than balanced point.. the column moment can't be push higher to reach balanced point moment before the yielding tension parts break apart, you think this is possible too?

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (mes7a)

is this a good way to think?

Mostly. I see two possible flaws in your logic:

1) you seem to assume that your existing columns were not properly designed to form plastic hinges when the structure was a four story building as originally planned. How do you know that is the case? If your columns are fine, you may well be penalizing your building -- and your sleep -- unnecessarily.

2) we design seismic plastic hinges to reach a certain capacity and then maintain that capacity while not resisting additional load. This is very different from a member or joint losing all load resisting capacity after the formation of plastic hinges. This needs to be kept in mind when considering the kinds of footing and column "failure" that may be acceptable.

You've asked for additional references several times now. And, in my estimation, the aspect of seismic design that you struggle with most is the general philosophy of capacity design. To that end, I recommend acquiring and digesting this reference: Link. It's one of the source documents that underpins much of our modern seismic design methodology. Normally, I would hesitate to suggest that someone read an entire, complex, out of date textbook in order to answer their eng-tips question. In your case, however, I know that you won't be deterred by the effort implied. And I'm confident that the exercise would answer a lot of the questions that nag at you.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
I have read that book before but ok I'll read it again for a week or so.

My point is simply this. Imagine you have a footing and the column is just half meter high and you connection a hydraulic bender to the column. It has zero axial load. Let's take the following case.



At zero axial load. Moment capacity is 180 ft.kips.
If you bend it with the hydraulic til it reaches 180 ft.kips, the rebars yield...

If you continue the hydraulic bender, can you push the moment to reach the balance point at 300 ft.kip? or would the yielding bars break halfway?

If it breaks halfway or before reaching the balance point.. then the footing flexural overstrength is dependent on the above.. isn't it.. ok.. this is my last question before I read the whole book again. Remember my designers told me their job is just to output reinforcement ratio in etabs.. they don't understand even the interaction diagram manual computations in any details or even plastic hinges.. when I told them about plastic hinges.. they said concrete is not made of plastic and just laugh at me. so I know I must seek others but need to know this basic before going on. Thank you.

RE: Horizontal Trussing Threshold for Roof Diaphragm

Quote (mes7a)

I have read that book before but ok I'll read it again for a week or so.

I'm simply trying to help. I'm not attempting to assign you redundant homework. Obviously, I have no way of knowing what you have and haven't read unless you compile a list for me.

Quote (mes7a)

If you continue the hydraulic bender, can you push the moment to reach the balance point at 300 ft.kip? or would the yielding bars break halfway?

In a properly detailed SMF column, neither of these things should happen. This seems to be one of the fundamental aspects of capacity design that may be eluding you.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

In a properly detailed SMF column, neither of these things should happen. This seems to be one of the fundamental aspects of capacity design that may be eluding you.

In a "properly detailed SMF column".. that's right.. but my column is not properly detailed SMF. The following pic shows why.



10 gallons of pure epoxy injected:



It is honeycombing in the column because of poor workmanship by incompetent contractors. The designers told me to inject 10 gallons of pure epoxy. In the thread http://www.eng-tips.com/viewthread.cfm?qid=341916, everybody here is convincing me the low modulus epoxy won't take much load and can't be used a repaired material. But the designers won't listen because they forgot the concept of modulus of elasticity or stress-strain so can't understand what BARetired is saying. In the following quote. BARetired is commenting I have loss of axial capacity of about 1000kN. BAretired said:

" Just a comment on your earlier calculations (shown in blue):

Now to compute for the load reduction carried by the epoxy filling. I'll use strain of 0.0005 or load in the elastic range.

from steel strain=0.0005, Modulus 29,000 ksi
stress = strain*modulus = 14500 psi or 99973.98 pascal 100MPa or 100*106 pascals.

from concrete strain 0.0005, Modulus 3604.996 ksi (let's call it 3600 ksi)
stress = strain*modulus = 1802.498 1800 psi or 12427.79 pascal 12.4 MPa

from epoxy strain 0.0005, Modulus 450 ksi
stress = strain*modulus = 225 psi or 1551.32 pascal 1.55 MPa

Column is 0.5x0.5m, the 0.2x0.5 section was replaced with epoxy, remaining 0.3x0.5 section with concrete. In other words, 33% 40% of section replaced by epoxy.

steel area of 12 20mm bars (for concrete section) = 0.003769 mm^2 3600mm2...in Canada, 20M bars have an area of 300mm2, could be different in Philippines
steel area of 8 20mm bars (for epoxy section) = 0.002513 2400 mm^2

For load carried by concrete section (0.3x0.5 of column) with 12 bars of 20m steel
P = Fc(Ag-As)+Fs(As) = 12427.29(0.146231) + 99973.98 (0.003769) = 2194.13 KN
12.4(300*500 - 3600) + 100(3600) = 2175 kN.

For load carried by the epoxy section (0.2x0.5 of column) with 8 bars of 20mm steel.
P = Fc(Ag-As)+Fs(As) = 1551.32 (0.097487) + 99973.98 (0.002513) = 402.4681 Kn
1.55(200*500 - 2400) + 100(2400) = 391 kN

For load carried by entirely concrete(0.5x0.5 of column) with 20 bars of 20mm steel
P = Fc(Ag-As)+Fs(As) = 12427.79(0.243718) + 99973.98(0.006282) = 3656.913 Kn
12.4(500*500 - 6000) + 100(6000) = 3625 kN

Loss of axial load due of the epoxy is
P(all concrete) - (P(concrete)+P(epoxy)) = 3656.913 - (2194.13+402.4681) = 1060.3149 KN
3625 - (2175 + 391) = 1059 kN

Note: The above calculation assumes uniform strain throughout the column. If a transformed section is used, the centroid of the combined section would shift toward the concrete portion. That would cause bending stress in addition to axial, so the condition is likely going to be worse than calculated."

Kootk. Have you seen construction this bad? In our country. Injecting epoxy is the normal even on column bases on 30 storey high rise structure that has 2 meters size honeycombing. BARetired said it's so dangerous.

It's good the honeycombing is in my tension side of the eccentric front column. If it is on the compression side.. it is so scary.. but reverse cyclic loading worries me. This is the main reason why I don't want to build the 4 storey.

Have you seen one building with 10 gallons of epoxy injected and massive carbon fiber installed and even pedestral retrofit done? Or have you seen worse.. lol..

This is the reason why Im learning what a properly detailed SMF column should be.. so asking if it is possible the yielding bar breaks before the moment reaches the balanced point.. etc. Thanks.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
Kootk... maybe can use your almost genius muldisciplinary knowledge one last time to bear on this problem that is only local to our country and yet completely ignored here in my country.

This concerns the holes in the column replaced with soft material (epoxy). This is our only solution in our country for most incompetent concrete pouring. The location is in red in the following building plan (base of center front column).



from side view the hole looks like this



The column involves is the 0.5x0.5 meter



I could have created a new thread of this but those who is not familiar with my structure may be confused as to its location.. for example BARetired who scrutinized it 2 years ago thought it was at midspan.

But the above shows it is actually in the level of the tie beam at front center column.

Both my contractor and designers are forcing me to accept it because they said it would perform like concrete.. but as the multidisciplinary team here emphasized. It functions like soft material.

My question is.. with the beams connected to it in above at at sides. Would the front column still bend via the hole with softer material epoxy (the bending stress BAretired describing below)?

This shows the front center column has perpendicular beams connected to it continuously from center (in the picture we haven't put the extra bars yet)



Let me quote what BAretired said about column with epoxy hole generally:

"If a column is hinged top and bottom and compressed with axial load P, the stress is uniform at every cross section, namely P/A.

If a rectangular notch is cut out of the left side of the column at mid-height, the centroid moves to the right at the notch. The axial load falls to the left of the centroid and the notched section will move right relative to the hinged ends.

If the notch had been filled with material with low E, the behavior will be similar, but the centroid will not shift as far so the filled notch will not move as far to the right as the unfilled notch because the fill is carrying some stress but not as much as the concrete.

You asked why the concrete would crush. You do not have a perfectly rectangular notch across the full depth of the column. What you have shown on your sketch looks like an irregular cavity. Some of the concrete extends to the left edge of the column and that is the part that would crush first.

If axial stress exceeds bending stress, there is no tension on any part of the cross section, simply variable compression with maximum value on the left and minimum on the right.

At a strain of 0.0005 and a depth of only 4" of epoxy, I would not expect enough strain in the column to cause the kind of problem described above. But if the column is carrying its full design load and the epoxy thickness is 12" instead of 4", I would have serious concerns about the safety of the structure."

Kootk.. strain of 0.0005 is in the elastic range. This is the reason I don't want to top off the full 4th storey because I'm not sure of the behavior and no one locally can scrutinize this level of details. So want to make it just be within the elastic range.

But during seismic reverse cyclic loading, the structure would move back and forth.. wouldn't this cause the bars in the holes with epoxy to yield because it's taking mainly the load from the softer epoxy? I've been thinking of this non-stop for the past 24 hours after reviewing that thread again. What is your experience of this.. what do you make of it? I did the following retrofit to fix the epoxy retrofit. But if the columns bends.. the concrete surrounding it may not restrain it (because the bars circling it may not be enough). Any other solution to maybe retrofit the 2 retrofits again?

retrofit rebars:



retrofit concrete:



Thanks a lot for all the multidisciplinary comments!

RE: Horizontal Trussing Threshold for Roof Diaphragm

Section 4 here has a procedure, it is using R=1 as part of the requirements. I must have been confusing R less than 1 with something I saw in AASHTO recently on connection forces under seismic.

https://www.wbdg.org/ccb/DOD/UFC/ufc_3_310_04.pdf

RE: Horizontal Trussing Threshold for Roof Diaphragm

I struggle to think of a suitable repair strategy mes7a. I agree, the construction quality is terrible and the existing repair quite unsuitable.

One could perhaps encase the entire first floor columns in new concrete and rebar. That would, of course, change many of the seismic parameters that we've been discussing.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

I struggle to think of a suitable repair strategy mes7a. I agree, the construction quality is terrible and the existing repair quite unsuitable.

One could perhaps encase the entire first floor columns in new concrete and rebar. That would, of course, change many of the seismic parameters that we've been discussing.

I have personally check and hammer each ground floor column. The others are ok. The reason the front center is problematic is because if you can see the layout below (arrow in red).. the tie beams are slant and the carpenters couldn't do proper formworks without leak and some rebars congestion there caused it.



I was hoping the column can act as gravity column only and the 8 other moment resisting columns are the main lateral resisting columns.. as you said to make the gravity column ride along with others.. Also someone mentioned in the thread 2 years ago that:

"Is this ONE column in a series or group of columns that contribute to the seismic resisting frame? If so, would it be possible that the remaining columns could support the lateral loading? This ONE column would only have to be retrofitted to support the vertical loads and detailed in a manner that it only 'go along for the ride' in a seismic event."

Btw.. our rebars here is 314 mm^2 for 20mm.. not 300mm^2 for 20M (which is in Canada). At bar strain of 0.0021.. the 8 rebars (at side only of the 20 rebars) have axial capacity of about 1000kN (i'm thinking how to convert it to moment capacity).. my building axial load above that column is only 700kN so the present section with concrete and epoxy and rebars have capacity of 2000kN. I'm still manually familiarizing myself with the interaction diagrams now reviewing what I learnt 2 years ago to see the composite behavior of epoxy and concrete. I'm doing this now because I'd add walls and light roof to third story and see how adding trusses can affect the load. My designers design the trusses don't worry. But they still keep telling me epoxy is the best repair because it is tested for 10,000 psi.. but when explaining to them the 10,000 psi only works when you compress it more than concrete (and strain incompatibility). They don't understand it anymore. I'm not making this up. But I can authorize them to approve any repair I want (such as manually removing each epoxy in cups and replacing with grouts) but the ground floor is already rented by tenants so can only do it if they leave.


RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
..

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)
http://www.eng-tips.com/viewthread.cfm?qid=395487

Kootk.. after learning more about interaction diagram and boundary conditions in columns from BAretired and how to derive the formulas. I think the safest is to retrofit braces frames, shear walls or even using stair shaft as lateral resisting system or turning it into gravity load only for the column with void at bottom.

Since you are familiar with my structure (before you forget them days or weeks from now). Please allow me to ask the following:

1. Since my bases are compromised (from voids filled with epoxy and insufficient flexural overstrength capacity in footing moment connections). The best way maybe to prevent rotations of the columns. Is it possible to brace the columns? with what? Like X frames between columns? How to put the X frames? Generally should such brace frame retrofit be connected to foundation.. it's difficult to dig the foundation again because of presence of tenants so shear wall may not be practical anymore.. I just want to prevent longitudinal movement (front to back.. see picture below) to avoid rotations of the columns that can stress the void in front (filled with epoxy) during extreme moment curvature). Of course I'll discuss this with my designers and they would design such system after getting some idea.. they don't have original ideas.

2. You mentioned above how stairwell shaft can be used as lateral resisting system. See:



In such system. Are the stairwell shaft connected to the columns or beams? In my layout.. it's not connected to middle columns so I wonder how you can consider it as main lateral resisting system.

3. You mentioned earlier that "In that case, the members not selected for primary lateral load resistance should be designed with sufficient deformation capacity that they can "ride along" with the designated lateral system without being compromised. I do this often.". For the column below:



Can it be made as column for gravity load only.. with the rest as moment resisting? But how do you make it have sufficient deformation capacity yet doesn't have lateral resisting effect.. you meant smaller column so it can't be converted (except by making the column smaller)?

Again. Thanks so much much. Don't worry we won't drag this thread to 200. Lol

RE: Horizontal Trussing Threshold for Roof Diaphragm

I've racked my brain over this Mes7a. However, I cannot think of a good, simple way to repair your building. My advice would be to somehow engage the services of a structural engineer that you trust and have them perform a thorough analysis of the situation.

I like to debate structural engineering theory -- a lot. If I challenge you on something, know that I'm doing so because I respect your opinion enough to either change it or adopt it.

RE: Horizontal Trussing Threshold for Roof Diaphragm

@KootK,

I haven't followed this thread all the way from the start but I am somewhat familiar with the problem. Do we know that a problem exists? It seems to me that the column in question will simply form a "soft spot" at the location of the epoxy filled void, a partial hinge so to speak. For gravity load only, moments will tend to redistribute to the top of the column and if that is not sufficient, to the beam which is applying the moment. We don't have any information on the analysis by the Engineer of Record so we don't really know how tightly he designed it. Perhaps he has a little spare capacity in other parts of the structure which could alleviate some stress redistribution. For seismic loads, perhaps the building can be rendered safe without reliance on this column at all.

I agree that the problem must be reviewed by a competent structural engineer. Why not ask the Engineer of Record to revisit the issue and provide an opinion. The building is occupied now so it is not something which can be put off indefinitely.

BA

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

I haven't followed this thread all the way from the start but I am somewhat familiar with the problem. Do we know that a problem exists? It seems to me that the column in question will simply form a "soft spot" at the location of the epoxy filled void, a partial hinge so to speak. For gravity load only, moments will tend to redistribute to the top of the column and if that is not sufficient, to the beam which is applying the moment. We don't have any information on the analysis by the Engineer of Record so we don't really know how tightly he designed it. Perhaps he has a little spare capacity in other parts of the structure which could alleviate some stress redistribution. For seismic loads, perhaps the building can be rendered safe without reliance on this column at all.

I agree that the problem must be reviewed by a competent structural engineer. Why not ask the Engineer of Record to revisit the issue and provide an opinion. The building is occupied now so it is not something which can be put off indefinitely.

The engineer on record said this problem is common in construction.. it occurs even on columns anywhere of 50-storey buildings. Once the epoxy company injected about 1 meters high of epoxy to a joint of a high rise. The designer on record said he forgot the concept of stress strain so just couldn't comprehend what I was saying. I'm not making this up. He said compressive test shows it is tested at 10,000 PSI. so no problem.. but when I said this stress occured at strain that is over 8 times that of concrete. He said he didn't want to discuss theoretical subject as he forgot them in school. When I visited other structural engineers. They said they also do it. So I have no one to turn to locally in my country. Hence seeking international multidisciplinary experts. Only left for me to let my voice be heard is to write an article explaining it all in the Structural Society newsletter. This is the only way for me to reach out to the contractor and designer who is foreign to concept of interaction diagram. He hired 20 newly graduates to operate the ETABs software which output the bars and that's what they gave to clients. When I tried to discuss with them. They said their only job is to output rebars numbers and it is up to the contractor to execute things properly. But the contractor is very incompetent.. so only thing left is to write a newsletter about this all so my voices would be heard.

About moment redistribution. What you mean by soft spot when you mentioned "It seems to me that the column in question will simply form a "soft spot" at the location of the epoxy filled void, a partial hinge so to speak. For gravity load only, moments will tend to redistribute to the top of the column and if that is not sufficient, to the beam which is applying the moment.". Won't the soft spot simply have maximum moment enough to put the bars in yield at the compressive bars and break the column. Why would it distribute the moments to top of column? I use a straw or tube to visualize what you all have been describing.. I cut a void in the straw and when I bend it.. the void just bends more.. there is no redistribution. I need to be familiar with this concept so if the newsletter readers used this same explanation. I can understand what he meant and can discuss it. Thanks a lot.

[Edit: maybe you meant the void section turn the column from fixed to pinned.. could that happen"]

RE: Horizontal Trussing Threshold for Roof Diaphragm

Yes, your edit is roughly correct. I don't believe it turns it into a complete pin, but under stress, the epoxy material will allow more rotation at that point, reducing the stiffness of the column. What this means is that the column is not as stiff as the designer assumed in his analysis, so the moment in the column from gravity loading is not as high as he calculated but other members meeting at the top of that column will be affected so the issue is not confined to a single member.

In the case of seismic loading, the story is a little different. The added flexibility of the lower column will not be a problem if the remaining columns are capable of resisting lateral forces by themselves. The partial hinge in this column will tend to prevent it from contributing significantly to ground accelerations which cause compression on the exterior face of this column. That means that the rest of the columns, stairwells etc. must be capable of resisting seismic forces without much contribution from the column with the epoxy filled void.

Perhaps the Engineer of Record could accommodate you by doing another computer run with a reduced stiffness at the void location. Seems to me to be the least he could do.

BA

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

Yes, your edit is roughly correct. I don't believe it turns it into a complete pin, but under stress, the epoxy material will allow more rotation at that point, reducing the stiffness of the column. What this means is that the column is not as stiff as the designer assumed in his analysis, so the moment in the column from gravity loading is not as high as he calculated but other members meeting at the top of that column will be affected so the issue is not confined to a single member.

In the case of seismic loading, the story is a little different. The added flexibility of the lower column will not be a problem if the remaining columns are capable of resisting lateral forces by themselves. The partial hinge in this column will tend to prevent it from contributing significantly to ground accelerations which cause compression on the exterior face of this column. That means that the rest of the columns, stairwells etc. must be capable of resisting seismic forces without much contribution from the column with the epoxy filled void.

Perhaps the Engineer of Record could accommodate you by doing another computer run with a reduced stiffness at the void location. Seems to me to be the least he could do.

The Engineer of Record is the president of the structural firm. Since they design 7 projects at a time.. half of them high-rise. The president didn't check all details.. for example.. he didn't know how many bars in my beam. But ok. I'd try to explain it to their design leader. Before the ground was covered with slabs and filling. I told the contractor to retrofit the column base with epoxy. The design leader looked at it and just ok it for peace of mind.. not necessarily to make the void section fixed. Have you designed any retrofit like the following?

picture shows epoxy section surrounded by rebars to confine the section.



concrete poured:



It won't make it fixed.. would it.. note there is only 2 stirrups (20mm) around the top of the section. Should the stirrup yields, the concrete confinement would no longer be effective?

In any case.. It looks like it may act like a pinned column base on a pedestral. Isn't it. Do you know of foreign firm who can run manual computations of my structure. In my country. Almost all use computer software so most doesn't compute manually anymore (it's labor intensive and time consuming for them).

RE: Horizontal Trussing Threshold for Roof Diaphragm

No, I can't recall ever designing a retrofit like yours. A concrete sleeve around the column. Vertical bars projecting above the sleeve, probably cut off flush after the pour. It is not clear how well they are attached to the foundation at the bottom; are they drilled and epoxied in?

2-20M stirrups? Are you sure? They look smaller than 20M to me.

The sleeve may provide some confinement, hard to say how much good it does. I would not rely on it to do very much.

Not many foreign engineers would attempt to analyze your structure manually. Computers are almost universally used nowadays to analyze structural frames. To engage a foreign engineer would likely be costly and communication would be difficult.

BA

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

No, I can't recall ever designing a retrofit like yours. A concrete sleeve around the column. Vertical bars projecting above the sleeve, probably cut off flush after the pour. It is not clear how well they are attached to the foundation at the bottom; are they drilled and epoxied in?

2-20M stirrups? Are you sure? They look smaller than 20M to me.

The sleeve may provide some confinement, hard to say how much good it does. I would not rely on it to do very much.

Yes. They are 20mm stirrups..the footing floor is drilled so the 20mm bars can be inserted. see the following pics showing them in more details:



Note the red arrow showing 20 means it's 20mm bars.



The idea being that if the orginal column crushes and yet the particles are confined inside the column, it can still resist much load.. it's like the concept of cans filled with marbles or stones.. any load imparted inside will be resisted by the materials inside via call confinement. So using the same concept. But it's not meant to make fixed.. but to confine possibly crush concrete.

Btw.. about this colume base fill with epoxy acting like pinned boundary condition. Well. In both sways of the seismic (front and reverse).. there will be one where the void will be in tension.. one where it will be in compression. The one where it will be in tension means the compressive side is fully concrete hence making the column take lateral resistance and can act fixed. while at another cyclic where the void become like compression, then it's like pinned.. so you have a vibrating condition where half cycle it is fixed half cycle it is pinned.. what would this be doing to the columns or any unwanted effect?

RE: Horizontal Trussing Threshold for Roof Diaphragm

The sleeve is more heavily reinforced than I originally thought. It should add considerable stiffness to the base of the column. For gravity loads, I don't believe there is a problem. For seismic loads, I'm not sure. It's hard to say how the combination of column and sleeve will behave. I think I would be inclined to leave it as it is rather than attempt further remedial measures.

BA

RE: Horizontal Trussing Threshold for Roof Diaphragm

(OP)

Quote:

The sleeve is more heavily reinforced than I originally thought. It should add considerable stiffness to the base of the column. For gravity loads, I don't believe there is a problem. For seismic loads, I'm not sure. It's hard to say how the combination of column and sleeve will behave. I think I would be inclined to leave it as it is rather than attempt further remedial measures.

BA

Not only that. On top of the sleeve are added 4 inches of slab with 10mm mesh poured right around the column (without any gap or membrane used), perhaps causing further fixity to the column. See:



My hope 2 years ago was to fixed the column right above slabs so moments would start above slabs avoiding the void from bending.

The 2 pcs of 20mm stirrups has confinement (shear-like) capacity of 2 pcs * 2 legs *314*470MPA = 600 kN. I wonder how much moments it can suppress. Any way to convert it to values of moments it can resist?

If the void really becomes fixed. What may happen (based on what you are saying about pinned) is that if
the sleeve and slabs break from the moments in the fixed base. Then the column converts to pinned?

Whatever, you are right that repairing it now means breaking the slabs and breaking the sleeve and the tenant won't allow so my remedial repair is to lessen seismic base shear by making the building lighter.

By the way. In our place. We don't use seismic membrane between column and ground slabs with foundation about 1.5 meters below. I'm worried about other columns. In your experience. Have you encountered the slabs (without seismic gap) compressing on the column causing shear failure of some kind? fortunately, the ground slabs are 3000 psi compared to columns 4000 psi.

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