CHS Moment Connection
CHS Moment Connection
(OP)
Easy one for somebody I reckon:
I have a chs which cantilevers out from a flat plate. It is fully welded to the plate with a fillet weld around its perimeter. How do I calculate the moment capacity of this connection? I have done 'what I think' is right, but it has now become the critical location which limits the load capacity so I need to check my method. I have a fixed weld size and the known weld transverse capacity.
I have a chs which cantilevers out from a flat plate. It is fully welded to the plate with a fillet weld around its perimeter. How do I calculate the moment capacity of this connection? I have done 'what I think' is right, but it has now become the critical location which limits the load capacity so I need to check my method. I have a fixed weld size and the known weld transverse capacity.






RE: CHS Moment Connection
RE: CHS Moment Connection
RE: CHS Moment Connection
RE: CHS Moment Connection
Moment arm = 2* centroid height of a half circle. determine your tensile force. see how much weld that requires, see if remaining weld works for shear.
This is just a rough calc that would let you know if it was going to govern. Some of the other designers here are much more adept at steel connection design than I.
I'll bet someone here might send you to "Design of welded structures" by Blodgett
RE: CHS Moment Connection
Have a copy of Blodgett and can't see this scenario, also don't have access to AISC examples (being UK based).
RE: CHS Moment Connection
Another option.
You can work out the shear and bending stresses in the CHS at the support and then multiply those values by the CHS thickness to get required weld capacities in tension and shear. Unless you really need to whittle the welds down, I'd just combine the max tension demand and the max shear demand, even though they'll occur at different locations. This solution is reasonable when the supporting plate is rigid.
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
RE: CHS Moment Connection
http://books.google.ca/books/about/Hollow_Structur...
RE: CHS Moment Connection
RE: CHS Moment Connection
That would be an elastic section modulus. You could also size the weld for the plastic moment capacity of the section using a plastic section modulus of the circular line.
RE: CHS Moment Connection
RE: CHS Moment Connection
RE: CHS Moment Connection
1) Often, I'll already have those numbers available from the design of the section and;
2) I find it to be a more intuitive way to incorporate the shear stresses into the weld design.
Six of one, half a dozen of the other.
This I'm curious about. I believe that Jayrod's solution implies plastic capacity as well. While steel sections are often ductile, welds themselves are much less so. Weld groups can't really deform plastically as far as I know. Do we trust that section yielding adjacent to the welds effectively shields the welds themselves from brittle failure under plastic strain conditions?
Can you elaborate on this Jed? We fillet weld baseplate for columns in tension all the time.
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
RE: CHS Moment Connection
RE: CHS Moment Connection
Thanks for the help guys, and for the links to the publications. I will get hold of those for future reference.
RE: CHS Moment Connection
To be honest, no one had ever shown me anything different. Now that I've got a new way to check welds on round sections I will definitely give 'er a go. It will be the thing I learned today.
RE: CHS Moment Connection
I agree that you shouldn't use the plastic section modulus for the weld for just an applied load where the attached section isn't also developing its plastic moment.
RE: CHS Moment Connection
RE: CHS Moment Connection
RE: CHS Moment Connection
This is just semantics, but the phrase "plastic weld capacity" and the like are a bit misleading for fillet welds in my opinion. Fillet welds themselves do not have reliable plastic capacities due to their lack of ductility. Rather, when members are expected to go plastic, welds must be designed to resist the applied plastic stress distribution while themselves remaining plastic. Semantics.
Having done some noodling on this now, I can think of two common examples of welds designed to deal with flexural plasticity in supported members:
1) AISC extended shear tab connections.
2) Moment connections in seismic frames.
Theses examples lead me to believe that welds should be designed to resist at least 1.25 Fy when the supported member is expected to go plastic. Essentially, the welds should be capacity designed using over strength of the supported member. Due to strain hardening etc, I would expect welds at the extremes of the section to apply tensile stresses to the welds in excess of Fy
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
RE: CHS Moment Connection
Side note. If anyone is looking to pick up books by Blodgett, they are very reasonable through Lincoln Electric.
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