Old steel truss design
Old steel truss design
(OP)
Hey everyone.
I have this steel roof truss comprised of double angles for all members with a 10' bay spacing for the bottom chord and verticals at the intermediate 5' marks for the top chords. This truss was designed and constructed in the late 60's under Canadian design standards.
I've checked a few of the components and it appears as though the original designer neglected combined action for the design of the top chord members. As far as I know your top chord must be supported at max 24" in order for you to neglect combined action or you must provide sufficient argument that the roof deck can transfer the load to the panel points.
When I run the existing members under the original loading (and using the applicable codes from the day for both loading determination and strength determination) for combined action they fail. Am I missing some provision?
I've used a slenderness of 0.9L/r where L is the panel length (5') and r of the 6x4x3/8" double angles (1.93 in) and the steel is 44ksi yield so I used 24.75ksi for allowable axial stress and 26.5ksi for allowable bending stress. Do these seem correct?
Granted it is not all of the members failing this way just the odd top chord member (I haven't got to checking the webs yet).
Would you consider continuous beam conditions for the top chord when analyzing for the moment? Something like wl^2/12?
I have this steel roof truss comprised of double angles for all members with a 10' bay spacing for the bottom chord and verticals at the intermediate 5' marks for the top chords. This truss was designed and constructed in the late 60's under Canadian design standards.
I've checked a few of the components and it appears as though the original designer neglected combined action for the design of the top chord members. As far as I know your top chord must be supported at max 24" in order for you to neglect combined action or you must provide sufficient argument that the roof deck can transfer the load to the panel points.
When I run the existing members under the original loading (and using the applicable codes from the day for both loading determination and strength determination) for combined action they fail. Am I missing some provision?
I've used a slenderness of 0.9L/r where L is the panel length (5') and r of the 6x4x3/8" double angles (1.93 in) and the steel is 44ksi yield so I used 24.75ksi for allowable axial stress and 26.5ksi for allowable bending stress. Do these seem correct?
Granted it is not all of the members failing this way just the odd top chord member (I haven't got to checking the webs yet).
Would you consider continuous beam conditions for the top chord when analyzing for the moment? Something like wl^2/12?






RE: Old steel truss design
Unless you need to report on the existing structure's original design, don't try. It is not easy as you bring too much baggage to the table.
If you really must try to get intl the validity of thr origin design (such as for a forensic report - been there, done that), well there are a variety of reason's that can cause this:
- Stop using the computer. The original designer didn't have one.
- Reduce the problem to simple node loading. That's likely a key point in your incompatible methods.
- Learn Graphic Statics. That's the most likely analysis tool.
- Pretend you only have the ability to know three significant digits, and occasionally guess a forth. Carry only four significant digits from every step to the next.
That's a start; if not, maybe I can help. Post the details of the truss and I'll let you know if it passes the old fashioned checks.
RE: Old steel truss design
It's more for my own volition to confirm the truss was designed correctly because the new loading I'm designing for is causing what I feel as excessive reinforcement in a few members. I.e. The total load added to the truss is in the range of 10% increase however I have members failing code checks into the 70% over range.
I can't post particulars quite this second but I'll ship you something later.
RE: Old steel truss design
I would say the original use of GS is your culprit.
RE: Old steel truss design
Off to la googlea
RE: Old steel truss design
I don't see what good it does here, jayrod seems to be concerned with the roof loading along the chord, not just at the nodes.
Michael.
"Science adjusts its views based on what's observed. Faith is the denial of observation so that belief can be preserved." ~ Tim Minchin
RE: Old steel truss design
RE: Old steel truss design
BA
RE: Old steel truss design
jayrod12,
I suggest a slide rule, or abacus...sorry, could not resist re a previous thread.
RE: Old steel truss design
You guys and your idiot boxes.... As I write this on my own idiot box. For shame!
RE: Old steel truss design
you mean one of these?
RE: Old steel truss design
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
RE: Old steel truss design
Parallel chord with a warren is correct. Johnny, tell the man what he has won.
It was analyzed using pinned joints. I plugged it into RISA for shits and giggles and played with the fixity of the joints. Fixing them makes matters much much worse.
RE: Old steel truss design
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
RE: Old steel truss design
RE: Old steel truss design
RE: Old steel truss design
compound truss...
Compound truss...
Compound Truss...
COMPOUND Truss...
COMPOUND TRUSS...
I had a couple beer at lunch. And I love compound trusses.
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
RE: Old steel truss design
We worry about far more than our predecessors, and as far as I am aware, combined actions were rarely applied before the 1970s. Many steel trusses I have seen where I had access to calcs (industrial buildings for the most part) had ONLY a graphics statics analysis as the basis of their design. That was it, that was all. After that they added some robust panel detailing able to take combined shear forces and occasionally couples.
I would be very keen to hear from some of the older members on this issue. I'm always keen to learn, particularly about older methods.
RE: Old steel truss design
@Jayrod: what have you modelled the unbraced lengths of the chords as? It's five feet or less for all buckling failure modes, yes?
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
RE: Old steel truss design
See attached for the truss makeup.
Note that i only sketched one half of the symmetrical truss. These trusses occur at 12.5' on-centre.
My members of concern are the 4th and 5th TC1 from the end.
Please expand on your compound truss theory if you could. I don't see where in this construction I could go that route. What would you consider the simple trusses that make up the compound truss?
RE: Old steel truss design
RE: Old steel truss design
RE: Old steel truss design
RE: Old steel truss design
www.SlideRuleEra.net
www.VacuumTubeEra.net
RE: Old steel truss design
RE: Old steel truss design
What is the particular mode of failure for the top chord Jayrod? Strong axis buckling? Overall section stresses?
I sketched some ideas in the attached PDF. Compound trussing probably isn't the cheapest way to go for this application, just the coolest way. If it was exposed architecture, I'd definitely consider it. Note that I didn't reinforce the right panels on your sketch. You'll get the idea.
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
RE: Old steel truss design
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
RE: Old steel truss design
Normally, there would have been open web steel joists spanning the distance between trusses. They were spaced at about 6' or 7' centers and the truss panel points would correspond to the joist spacing thereby eliminating bending in the top chord. This was done so that the steel deck could be 1.5" deep 22 ga and the flutes ran parallel to the truss.
It was the policy in our office to use Horizontally Braced Frames (HBF) spaced at about the quarter points of the trusses. These were rigid frames of the same depth as the trusses which kept the trusses vertical, similar to bridging used on open web steel joists.
Offhand, I do not recall designing a steel truss where the load was uniformly distributed along the chord, but if that situation had arisen, I would have taken into account bending using a coefficient of, perhaps wL2/16 for positive and negative moments in the central region of the truss and a little more near the ends.
In Toronto, as I recall, the plan checkers required a graphical stress analysis to be shown on the drawing for trusses. To keep them happy, we included a stress diagram but I always analyzed each truss using recognized non-graphical methods. IMO, analyzing a parallel chord truss by hand methods was much easier and less time consuming than preparing the stress diagram.
BA
RE: Old steel truss design
Anyone else able to add to the historical record?
RE: Old steel truss design
The greatest trick that bond stress ever pulled was convincing the world it didn't exist.
RE: Old steel truss design
Since everything was done with allowable loads, you could include all of the governing loads (DL, LL, SL, etc.) and check the governing condition. Most roofs were built-up roofs with gravel, so we never really worried about wind uplift, etc.
Through the '70's, it was common to include a Maxwell Stress Diagram on the design drawings so that the fabricator could graphically get the member forces. My guess is this is the graphic analysis noted above.
Were we overly simplistic. Sure! But trusses had been successfully designed that way for decades. Railroad and highway bridges, as well as buildings. Were we missing some secondary stresses? Probably, but we met the standards that were set by AISC, etc.
In my later years, I would often have to check an existing truss for new/added loads or load points. The easiest way was to plug them into a computer. Often the continuity would show overstresses in the chords. I could remove the self weight and make sure the joint loads included the total dead load. Then checking the chords as completely pinned or continuous to the points of chord splices would often give adequate results.
The lateral bracing of the trusses was done with what we called "sway frames". They included struts at the top and bottom chord elevations along with chevron or V bracing. They were placed at the third or quarter points as required to make satisfy the out-of-plane conditions for the trusses.
Most of us lived in 2-D worlds. Worked the plans and the respective cross-sections and elevations with 2-D structural solutions. I realize that it is now a 3-D world and everyone wants to model everything in one model and then hit Enter and see the results. Then whenever something is > 1.000 (and turns red on the model), regardless of how close to OK it is, they just go bonkers.
I often joked that 2D and 2D and 2D equals 6D which is > 3D. I know you aren't going back to the 2D world, but things were really so much simpler then and I feel the added complexities introduced to the Codes are because of what computers can do for the designer, but it seems to be an out-of-control spiral.
And while I am ranting, don't get me started on FEA. It is an approximate method and seeing some of the plate elements turn red does not necessarily mean failure. It may mean that your model needs refining, or you need to take a closer look at those isolated conditions with other methods of design and analysis.
gjc
RE: Old steel truss design
BA
RE: Old steel truss design
What is your confortable level of overstress? Depending on the analysis method, the application, the material at hand, and what I have made in assumptions, I am often happy at 1.02 and 1.05.
RE: Old steel truss design
So I would have no problem with 1.05. Some of my latest supervisors could not accept ANY overstress. We always joked about one in particular "seeing red".
gjc
RE: Old steel truss design
i think the 24" may be from the current SJI specification for steel joists based upon testing for their manufacturers, at least, that's the only place I've heard that number in this context.
BA: Normally, there would have been open web steel joists spanning the distance between trusses. They were spaced at about 6' or 7' centers and the truss panel points would correspond to the joist spacing thereby eliminating bending in the top chord. This was done so that the steel deck could be 1.5" deep 22 ga and the flutes ran parallel to the truss.
i agree with BA on this, but/and is this the way your system is? In other words, is there a mechanism for the direct introduction of bending into your top chords?... or are all loads introduced at panel points?
OP: Would you consider continuous beam conditions for the top chord when analyzing for the moment? Something like wl^2/12?
i might. But it seems to me that if it is a system as BA described above, alternate panels may deflect upwards and other panels deflect downwards creating a different KL/r and hence a different wl^2/?. In other words, how is the top chord braced/restrained?
double angles are fun, especially if z-axis buckling is overlooked.
RE: Old steel truss design
The method of failure that I get for even strictly axial load (neglecting bending) is flexural buckling.
The webs and chords are all connected by ~1/2"(12mm) gussets.
These trusses do have some form of uniform load as the roof deck is 3" 20ga decking that bear directly on these trusses.
There is bottom chord bridging at every second bottom chord joint and vertical cross braces on the bridging lines every 5th truss. Top chord bracing 6 bays out of 44 evenly spaced.
The 24" that I mention existed in CSA-S16-61.
Maybe someone can explain this to me. In my RISA model, when I run the truss (loading at joints only, all members pinned) using the AISC 14th edition ASD and allowable loading I get the 5th and 6th top chord member fail at 1.06 (I could live with that probably).
However when I run the same truss using the CSA S16-09 (my ideal code check) and factored loading the member fails by 150%. When I hand checked the same member using the CSA S16-09 and factored loading I get a code check of 117% in flexural buckling. I've attached the risa model in case any of you want to take a crack at it.