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Very Non-Standard Connection, Would like an Opinion.

Very Non-Standard Connection, Would like an Opinion.

Very Non-Standard Connection, Would like an Opinion.

(OP)
Greetings, I am currently designing a moment connection between an HSS column and an HSS beam, I am forced into this configurations and although I would very much prefer to use W-shapes it is out of the question. Initially I tried to simply weld the beam to the column, but this lead to a myriad of problems such as plastification of the column wall, punching of the column wall and such. Currently I am trying to use stiffening plates to keep the wall of the column from deforming. However I know of no hand calculation to design or even estimate the design of this connection, this led me to rely on a FEA program (SAP2000), which I very much dread since I have to rely completely on the computer. What I did is model the connection without the siffeners and check that its behavior is as expected (miserable failure of the the connection to behave as fixed) and then I proceeded with my stiffener plates scheme. The general behavior of the connection is as expected, the HSS walls are no longer deforming as the plates give them considerable resistance. I am seeing very high stresses near the corners of the plates.. But this is a linear FEM Analysis so no stress redistribution after yielding is being accounted for. I will attach Images and printouts of the connection and would very much like to hear your input, If anyone can suggest a method for hand calculation I would greatly appreciate it.

RE: Very Non-Standard Connection, Would like an Opinion.

Looks like a good idea.

What about a simple calculation for direct shear in the plates and the stress in the welds, due to load running in the same direction of weld along the side of the HSS column?

The definition of a structural engineer: overdesign by a factor of 1.999, instead of the usual 2.

RE: Very Non-Standard Connection, Would like an Opinion.

How about using an end plate larger than the width of the column, welded to the HSS beam (all around)in shop, and then field welded to the HSS column (all around). This will eliminate the effects on the HSS column walls and give you a symmetrical weld which will not cause any eccentricities and will be enough weld to resist loads.

RE: Very Non-Standard Connection, Would like an Opinion.

(OP)
@AELLC: As usual you have been most helpful, it occurs to me know that for the purpose of sizing the plates i could treat the connection as if the plates were gusset plates. Having this in mind I could check the plate for: Block shear due to the welds, Gross Section Shear Yielding, Gross Section yielding (using witmore width starting from where the beam first meets the plate, this would also give me a rational way of determining how long the plate should extend over the beam, based on the 30 degrees used for witmore). In the compression plate I would say that buckling is restricted due to the webs of the beam which are in contact and welded to the plate. I do wonder how to account for the beam webs welded to the plates when calculating yielding of the plates, as they must add resistance to the plate. I will definitely use your recommendation for the weld stress as its the only reliable way of designing the welds I can think of.

@civeng: Thanks, this is a very good idea. Another connection I must design has two beams meeting the column from opposite sides, this will most likely require a solution similar to the one you propose; a sort of "ring" plate around the column.

RE: Very Non-Standard Connection, Would like an Opinion.

Are you familiar with this? - it was very useful back when did steel design.

http://www.amazon.com/Design-Welded-Structures-Ome...

The definition of a structural engineer: overdesign by a factor of 1.999, instead of the usual 2.

RE: Very Non-Standard Connection, Would like an Opinion.

Actually, this is the best place to buy -- worth every penny.

http://www.jflfoundation.com/ProductDetails.asp?Pr...

The definition of a structural engineer: overdesign by a factor of 1.999, instead of the usual 2.

RE: Very Non-Standard Connection, Would like an Opinion.

(OP)
Thanks I just purchased it

RE: Very Non-Standard Connection, Would like an Opinion.

Would using 50 ksi steel help? The tubes are most likely 46 ksi, right?

RE: Very Non-Standard Connection, Would like an Opinion.

(OP)
Indeed.. the tubes are 46ksi, however I dont think that plates of the same steel or superior are available commercially here. Best thing I could do is use cuttings from the beam's web as plates. But they go very low, only up to 0.9cm while the A-36 plate I'm considering is 2.54cm (1"). Now that I have a hand calculation method for the plate Im getting values much smaller (1cm plate) than that of what would seem appropriate with the FEA, as in the FEA I can see stresses going above the yield value in much of the A-36 plate when using the 1cm plate (enough lead me to believe that redistribution would not be enough). One thing that does holds true in the FEA is the witmore width, I get a very similar V shape on the stress distribution. Honestly I would trust the provisions in the AISC specs over a FEA any day of the week; but if the FEA suggest a more conservative approach I would prefer to err on the side of safety.

I believe that by extending the plate all the way to the lenght of the column sidewalls I will be effectively avoiding any of the failure modes of the HSS walls (except shear due to the weldings). Any thoughts on this?

Thanks, this forum and its users are a tremendous help.

RE: Very Non-Standard Connection, Would like an Opinion.

I usually use the min overlap of pl on beam to be the width of the beam. Using the witmore approach is ok and also checking for shear lag as per AISC. I second the recommendation of civ in making the pl a ring on the col. The weld of the pl to the bm will be a partial penetration flare-bevel weld whose effectiveness is 5/8's of depth of weld. You are on the right track with your concept and also in questioning and confirming the model output. Not knowing the magnitude of loads or member sizes, I am curious why you had to add an extra seat @ the bottom of the bm, unless to take the verical shear, if so, I would add a closure pl @ end of the bm. In HSS design, I often find that the connections often drive the size of the member that I choose to facilitate a reasonable and practical conn.

RE: Very Non-Standard Connection, Would like an Opinion.

(OP)
Thanks for your reply, I followed the advice of "wrapping" the plate around the column and it is actually the only way I can develop enough resistance to the tension force in the top plate. I am using 3/4" fillet welds between the plates and the HSS, on the top and bottom surface.. I chose 3/4" because my understanding is that this is the maximum allowed for A-36 steel using 70ksi welding. I have a question, I envisioned the weld as a simple fillet between the plate and the column wall; why would I need a partial penetration weld?. I don't know if you are referring to the bottom plate or the second plate, the bottom plate I added because the compression force coming from the haunch is enough to punch in the column wall; the middle plate I added only for redundancy and I am considering removing it. I usually do the same for connection of HSS trusses.. but in my situation I have to use a relatively large HSS column (20x20cm) and all the beam sections that I could choose from where either 26x9 or 30x10, in both cases the flanges are slim enough to cause punching and plastification of the column.. so i find myself in need of this stiffeners.
Im trying to have the connection take a moment of 16000 kgf.m or 116 kip.feet, this beam runs 40 feet between columns but at 20 feet it rests over another beam that runs perpendicular to this one.

Here no one construct using this stiffeners (no one actually calculates the connections), so my client may think I am crazy when I give him the drawings, But theres no way this connection would take any moment without such stiffeners.. And I know for a fact that others assume this things to be fixed since they put haunches at the ends of the beams.. why else would you use haunches if not to increase the moment capacity at the the fixed end of the beam? My guess is that in reality this connections act as pinned and all that moment goes to the middle of the beam.. the only thing saving this guys is the 1.2DL and 1.6LL.

RE: Very Non-Standard Connection, Would like an Opinion.

In block shear, I would not include tension allowance.

The definition of a structural engineer: overdesign by a factor of 1.999, instead of the usual 2.

RE: Very Non-Standard Connection, Would like an Opinion.

(OP)
Do you mean disregarding the tension plane and using only the shear planes? Even by neglecting this I still have enough capacity, the controlling limit state is shear yielding at the whitmore width.

My hand calcs give a 45000 kgf capacity for a 1cm thick plate. Which I would be quite happy with (I need only about 35000). I try to test this on the FEM and Im getting stresses as shown on the attached picture. The deep blue color are stresses above the yield stress of the plate. I would like to somehow reconcile this results with my hand calcs... It seemed to me that when I use whitmore width for the plate yielding limit states I am assuming that all the yielding occurs along the whithmore width.. In the FEM This is not happening.. I have the "triangle" resulting from the whitmore width and the lines starting at 30 degree.. however the inside of the triangle is not yielding, while its perimeter is.. When we do whithmore width.. are we assuming that all the high stress is being redistributed from the point of force aplication "down" trought the plate?. The stresses in the picture occur when I apply a tension force of 40000 kgf. I would attempt a non-linear material analysis but this takes days to run. Basically I am just trying to validate my hand calcs so I can get rid of the FEM, which honestly I quite dislike using; but I dont want to ignore some important limit state by mistake.

RE: Very Non-Standard Connection, Would like an Opinion.

When you get this whole procedure down, you should automate it with Excel, then you won't need to run it on FEA.

The definition of a structural engineer: overdesign by a factor of 1.999, instead of the usual 2.

RE: Very Non-Standard Connection, Would like an Opinion.

(OP)
Oh I do everything in excel, I call hand calc's everything I do with a spreadsheet... I have a spreadsheet or maple (similar to mathcad) file for pretty much every situation I've had to face in the past. It saves so much time.. but lately I seem to be facing completely different stuff in every different project. My ongoing hobby is to get all the spreadsheets to draw data from one access file, I've manage to do this with section properties and rebars; for example in my current HSS capacity spreadsheet I can push a button and then choose from a list of sections pulled from the access file. If I want to add a new section I just open the access file and add a line with the data then the section is available in all the spreadsheets that get data from this database. My long term goal is to be able to have a database keeping information for each structure I work on.. say the nodes, the members and the moment and shears.. so I can get this values into each spreadsheet design and then save the design data back into the database.. But I never have time.

RE: Very Non-Standard Connection, Would like an Opinion.

(OP)
I think I found my error in the model.. I was applying the loads at only a few points at the end of the plate... Now I am testing different configurations of loads on the plate.. The one that I believe is closer to reality is where I apply the Tension force coming from the beam's top flange all along the weld line that joins the top flange of the beam with the plate. I will attach a pdf showing the force application and resulting von mises stress distribution in case anyone is willing to give it a look and comment.

RE: Very Non-Standard Connection, Would like an Opinion.

IngDod:
The FEA looks about like you would expect it to look. And, your FEA printout/picture is telling you exactly what to do. Show us a couple dimensioned views of that connection, what size are the two members? What are the side pls. made of, that make up the hunch? Are they full height, top of beam tube to the bottom sloped surface, a bottom flg. in compression? You see that you can eliminate the middle stiff. pl., it doesn’t do much, except complicate the fab’ing. The bot. stiff. pl. might be placed at the same slope as the bot. of the hunch. And, you might try reshaping the top stiff. pl. so it takes the top tension out to the sides of the column tube, as a “Y” shaped (forked) tension field, not in bending across the stiff. pl. and the face of the column tube. Maybe that top stiff. pl. has an elliptically shaped cut-out, with the minor (major?) axis being the width of the col. tube, and running half way out the length of the hunch over the top of the beam tube to a half circle termination. This completely eliminates top stiff. pl. mat’l. in the yellow area of your latest picture. The outer edges of this top stiff. pl. should taper out on top of the beam tube about the length of the hunch, maybe a little further, as you’ve shown in your latest sketches. Try running something like this on the FEA and see what it looks like stress-wise. It should completely change the blue stress picture around the column corners. You will always see high stresses in your FEA, at corners, the multi-axial stresses kill you. And, we know this from our Theory of Elasticity study. We’ve known for a long time welding into and around corners can cause problems. But, a second angle of attack (a consideration), regarding the high stress areas in your FEA pictures is that, as long as a small region in a detail is well confined, by surrounding material or weld, a little yielding is usually not a big problem. There is still compatibility and just some redistribution of stresses/forces to the adjacent material. All of this needs a little more study, once we know the proportions/sizes of the connection and the members, and the loads and moments. I have to reread your last couple posts, and think on them a bit.

The book that AELLC suggested would be well worth your while, and Lincoln has several other good books on welding and weld design too, several of them by Omer W. Blodgett. “Design of Weldments” is one of them, with some different material in it. You have some really difficult welding conditions on this detail, and the FEA doesn’t show the potential problems in the welds. But some of them are at the same locations as the high stresses shown on the FEA.

RE: Very Non-Standard Connection, Would like an Opinion.

(OP)
Thanks, I tried the FEA with the shape you described. Honestly I didn't quite understand the part about the shape close to the column face. I tapered the top plate as in one of my previous sketches.. It doesn't seem to have that much of an effect, other than saving material perhaps. I am adding a pdf wih the new results as well as the dimensions and materials of the connection. The moment im designing for is 16000 kgf.m or 116 kip.ft, the baseplate for this results have a 2" thickness. All other dimensions are on the pdf. I removed the middle plate, there's no appreciable effect on the connection. Thanks for your help.

RE: Very Non-Standard Connection, Would like an Opinion.

Try to find a copy of Packer and Henderson's text, "Hollow Structural Section Connections and Trusses" one of the best HSS connection design texts out. Also CIDECT has some excellent publications.

Dik

RE: Very Non-Standard Connection, Would like an Opinion.

IngDod:
Regarding the top stiff. pl.... The ears on my stiff. pl. (along the col. sides) would be 5-6cm wide, as your’s are, but they would be 25-26cm long (not your 22cm) before they started tapering into the 9cm beam tube top flg. I would increase your 18cm length to 24-30cm, further out onto the beam tube. At the column face, place the center of a compass at the face of the col. and at the center line of the beam and draw a half circle cut-out, the circle dia. is 20cm, the col. width. Except, I would like the cut-out to be elliptical in shape (minor dia. 20cm, major dia. 30cm), going out onto the beam a little further than 10cm. My top stiff. pl. is “Y” shaped. The large corner blue areas will go away, I’ll bet. I agree that tapered or square won’t make much difference on the stress flow in the top stiff. pl. but it is probably being cut out of a larger pl. by automatic cutting equip., so why not clean it up? Eliminate the middle stiff. pl. it just complicates the assembly. The bot. stiff. pl. should be tipped up about 30̊, made longer, and becomes the bot. flg. on the hunch, but it is cut square to fit the col. shape. Alternatively, the hunch could be cut from a piece of the beam tubing; then the stiff. pl. can be shorter, but I’d still tip it up to match the hunch slope.

Don’t weld around corners, stop welds on the flats of the tubes. The longitudinal welds btwn. the stiff. pls. and the corners of the beam tubes are really difficult welds to do well. Yes, you should include the load inputs from both the weld across the top of the beam tube and the two longitudinal welds btwn. the stiff. pl. and the corners of the beam tube, but do not try to connect them by welding around the corner of the pl. at the tube corners.

RE: Very Non-Standard Connection, Would like an Opinion.

IngDog - You may have already considered the following:

Make sure the architects won't have heart attacks when they find out the connection projects 50mm (2") beyond the face of the column.

Are the 3/4" fillet weld on both sides? Consider partial penetration of full penetration welds.

With such big welds, it may be worth checking for base metal rupture of the column (In the US, 5/8" is the maximum wall thickness for HSS8x8).

Check the beam for the concentrated load from the haunch. Stiffeners may be a good idea even if the beam can handle load. It could be informative to extend the beam past the haunch in the FEA model.

RE: Very Non-Standard Connection, Would like an Opinion.

(OP)
@dik: Thanks, I already have this book and most of the CIDECT guides. They are a very good resource.. but they are remarkably light on hss to hss moment connections.. they do say that directly welding them is a big no-no for developing fixity unless the sections are of similar width.

@dhengr: I understand now, well honestly I rather leave them as squares cause most likely this things will be cut using a hand-held torch (I live in the third world).. so any fancy tapering could come back and bite me in the rear. I will most definitely keep your suggestions about welding in mind.

@wannabeSE: Im pretty sure he will, but he is the one who wanted to have a 40feet clear span while supporting a steel deck. (this beam is supported on another beam at 20 feet.. just don't want to scare anyone thinking this thing spans 40feet on its own). I will have a pretty tough time designing the welds. I will keep everything you told me in consideration when designing the welds. I had the beam going past the haunch before.. I did see some stresses going over the limit but only in the vicinity of the haunch (no more than 2cm i would say) the rest of the web was fine, I will have to remodel and check again.. the problem is that I am applying the moment to the connection by using forces on the joint of the top web, directly above the haunch, so this will cause an artificial high stress in the haunch.. I will have to do a separate model for this.

I finally modeled the whole connection, it has a very large beam framing on one face and two more beams framing perpendicular to this one. Since I had to greatly expand the plates to take the other beams Im seeing much better stress distributions and my plate is only 1cm thick. However I am having a hard time finding out the proper way to get the design forces for the welds.. I mean I can probably get it with the FEA, but I would like to be able to do some ballpark estimating by hand. Would summing the tension forces (as vectors) coming from all three beams top flanges and using the resultant be an appropriate way? I am attaching my latest results, please comment.. this easily the weirdest connection I have ever designed..... Now I have to do two more.

RE: Very Non-Standard Connection, Would like an Opinion.

(OP)
@dhengr: I will definitely try to angle the bottom plate with the haunch, I will look much better and be more effective since its aligned with the load.

I have a question.. If I size the welds to be enough to develop the rupture stress of the base metals, would this not preclude a failure of the welds? I mean, the hss or the plates would have to tear before the weld fails.

RE: Very Non-Standard Connection, Would like an Opinion.

Since the Arch is being a little difficult, can you CJP a solid block of steel at the connection point and weld to your hearts content?

Dik

RE: Very Non-Standard Connection, Would like an Opinion.

(OP)
Well i was just trying to be funny... This is a steel deck structure so most of the connections will be hidden behind a drywall ceiling... so the aesthetic of the connection are not very important. I will prefer CJP.. but I am concerned that the hss walls are very thin.. only 0.55cm... Also a concern of mine is the skill of the welder, here welders are not certified as in the US. Some are very good, some wont know the difference between a fillet and a CJP.. I was speaking to one the other day and I asked him.. how you make the weld a certain thickness? he told me... "I Just make many passes".

RE: Very Non-Standard Connection, Would like an Opinion.

The architect may be concerned if the connection interferes with the framing for exterior cladding. Or, if it conflicts with studs for walls used to conceal the columns.

RE: Very Non-Standard Connection, Would like an Opinion.

These are really large loads for this conn....I get 60kips(approx) for T&C....60/.707=84.8 kips @ the haunch...how are you going to transfer this to the col?....the majority of the T & C loads will be reacted by the sidewalls of the col so the center portion of these pl's will not see any appreciable load...the load basically stays in the outer region of these pl's...I attached another concept that may be viable but you may run into problems with getting at some of the welds...another option is to play around with differnt width bms inorder to accomodate some of these welds....also this is imparting significant shear loads into the col sidewalls and when combined with any other existing loads in the col may need to be checked.

RE: Very Non-Standard Connection, Would like an Opinion.

(OP)
@SAIL3: I greatly appreciate you taking the time to help me with this.

The connection in your sketch would be ideal, I believe the intention with this is to have the gusset transfer the load directly to the plate?. In my design the haunch loads the wall and then the wall loads the plate I believe (the wall tries to punch in and the plate restrains it). Yes, my intention with this stiffening plates is to allow the T and C force to travel from the front wall (the one to which the beam is framing) to the sidewalls without punching or deforming the front wall. The way I first envisioned the load path is: The beam flanges push and pull on the HSS walls, the walls then pull and push at the plates, which keep the walls in place and distributes the load to the sidewalls. Although I believe that by welding the flanges to the whole connected length of the beam I am transferring the load to the plates and then the plates transfer to the side-walls; however this is not the case with the bottom plate as I have it now, as the haunch bears on the wall without being connected to the bottom plate. I will definitely try to see if your connection detail is viable, I will have to talk with the welder.

This columns are loaded mostly by the moment coming from the beams.. their DC ratio is around 0.2 Compression and 0.8 Bending. I did not manually check interaction of axial and moment stresses with the shear coming from the T and C forces.. this is something I will have to do.

RE: Very Non-Standard Connection, Would like an Opinion.

Due to the thin walls of the sections, moment connections in HSS will begin to look more like space frame connections as the moment gets higher. The stiffened connection will begin to look like a socket that the tube plugs in to.

Here is just an example I googled of a wide flange beam in an HSS column (see figure 3). You can see that the entire HSS section is wrapped to ensure the integrity of its shape...

http://www.atlasconnection.com/profiles/blogs/hss-...

I think the design you show with stiffeners on the top and bottom is a holdover from wide flange theory where most of the moment goes through the flanges. This will not be the case I think with HSS and there will always be tearing and distortion.

The most important thing to reaching the full moment capacity of the section is to ensure the integrity of the entire HSS section shape. I think you will find that using gusset plates and stiffeners to create what is effectively a welded socket is the best way to do this even though it appears redundant and impractical in wide flange world. The benefit is that the connections should be very straight forward to design.

I have sketched what I think the connection might look like if the beam and column are the same width. If you want the column wider than the beam then the gussets and stiffeners will need to transfer the load not the HSS.


RE: Very Non-Standard Connection, Would like an Opinion.

another item to check.....assuming the t of the col wall is 0.55cm=.22 in......T=C=60kips...that is 60/2=30 kips per sidewall of col.....shear area=(8-3x.22).22=1.62in^2.....shear stress=30/1.62=18.6ksi....also I would check for shear panel buckling in the col sidewall between the top and bottom rings...I would either increase the wall thickeness or if that is not possible add a flat pl to the sidewalls in that area....
charlie brought up some good points and referenced an interesting site...

RE: Very Non-Standard Connection, Would like an Opinion.

(OP)
@Charlie: Thanks for the link and for the sketch, I had read that website during my research. The connection you propose is exactly the same as the one proposed by CIDECT to develop a rigid moment connections in Vierendeel trusses, I tried that but there's no way I can get a beam wide enough to match the column and the architect would undoubtedly object to the haunches protruding from the steel deck. I had a conversation with the manufacturer of the HSS in my country, they tell me that in this cases the ring devices I am trying to develop have been used in the past, they told me: "The plates act as a diaphragm by distributing the stress to the walls of the tube", they also told me I should keep the thickness of the plate at no more than twice the thickness of the hss for stress distribution reasons.

@SAIL3: I have not managed to get to this part as I was just requested to modify the structure (architectural reasons), this presents me with the opportunity of increasing the beam size and getting rid of the haunches, which actually lowers the moment on the connection by about 15%, And would simplify the connection. Im afraid that's as thick as the walls of these tubes get, the steel is Fy=50ksi and Fu=62ksi so it might hold, welding a diagonal plate from top ring to bottom ring could help in this (similar to what is done with W-Shapes) and I will also check the panel buckling.

Again thanks to everyone who has contributed to this thread.

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