Who’s the design checker? I would tell them to stop wasting my time. The only moment it’s throwing into the wall panel is the eccentricity of the beam reaction which wouldn’t even exceed the cracking moment of the panel.

If that’s the detail the precaster and steel fabricator have been using for years then there would be no need to change it now. I generally to a similar detail with a seated angle bearing connection fixed to the panel with cast-in ferrules.

I don't see a problem with the seat concept, as long as it is not at a panel joint. The checker may be concerned about the position of the girt, and the restraint by the girt as the rafter rotates.

He's concerned that the connection will not act as a pin connection and that it will induce a moment on the panel which could be dangerous.

With the girt located near the top of the rafter, I would share his concern. You might find that the trubolts nearest the rafter are failed by the rotation.

This is a fairly standard detail in the precast industry.
The girt you refer to Hokie is in fact an eaves beam used to support the panels at the top and also forms the bottom chord for roof bracing.

Are you suggesting the channel should be placed lower to reduce the rotating moment ?

Just a couple of questions, as I know Down Under you use a few systems that are not as popular in the USA.

What is the horizontal channel girt for? I know it braces the top of the roof beam against rotation. But you said it braces the panel, which makes me also wonder is the roof deck not a diaphragm? It would be more common in the USA to have a shelf angle that supports the roof deck, which is the diaphragm, and this braces the top of the roof panel and transfers in-plane and out-of-plane loads.

That aside, the bottom angle seat will put a moment into the panel, usually I would use about 2" of eccentricity for that connection and design the embed plate for that amount. However, your top rebar will take that up in tension so that looks OK.

The top channel girt COULD transfer some force into the panel when the roof beam is loaded and the top rotates slightly. However, I think using a bolted connection there prevents much restraint, and I think you could provide slotted holes that would allow the top of the beam to rotate (in-plane) but would provide lateral-torsional bracing.

Why the need for the 200PFC at all?? What is it for?

I'm located in the U.S. For roof beams to panels, we typically use a double-angle shear connection with horizontally slotted holes welded to an embedded plate in the panel. You already have the embedded plate in your detail, so you can easily do that. Either way, you only need to consider the load eccentricity of the connection for moment in the panel from the beam.

civeng80,
Yes, placing the channel lower would reduce the tendency for the rafter rotation to place the trubolts in tension. But I realize that you have to coordinate that position with the bracing plane. Another reason for lowering the channel is that, at the lower end of the roof, it often interferes less with the gutter, sumps, and downpipes.

a2mfk and ztengguy,
The channel, as civeng80 said, is to support the panels at the top, and also to serve as a chord for a roof bracing system. We don't tend to use deck diaphragms here. We use steel roofing which is fastened either through the crowns or with a concealed fastener system, and the roofing is considered to have no diaphragm capacity. Rather, a horizontal truss system is employed.

steellion,
Double angle shear connections are another thing not typically used in Australia. Typical shear connections are end plates and fin plates.

Spot on Hokie66 and thanks all for responses so far.

The guttering and sump and downpipes are not really a problem, but bracing further down would possibly mean putting in CHS struts/ties and not rely on the purlins for bracing. This is Ok as I dont really like using purlins (even double purlins) for bracing members.

The point is what is a safe moment arm from the angle seat to the girt cleat so that the moment transfered to the panel is safe.

Been thinking maybe a beam support with a spring with a stiffness equivalent to 2 cantilevers about .5m long (the distance from the centre of the beam to the first trubolt) woud give me an idea of the moment transfer to the panel, but this is statically indeterminate.

Even a ball park reasonable figure for the moment transfer to panel would be good.

Has anyone got a solution for this problem?

I agree with you about using the purlins as struts. I know some people do it, but I never have.

The further you move the trubolts away from the rafter, the better.

Thanks Hokie.

Any ideas about how to actually estimate the moment transfered to the panel with this (Semi rigid) connection?

No, I don't. Sorry.

Fair enough.

Thanks again.

civeng80,

I have a detail very similar except one minor difference. Instead of the M16 Trubolt @ 1000 cts What you can get is a 'clip' type element which is fixed to the precast panel via a ferrule and it is not fixed to the PFC, however the flange of the PFC sits between this clip and the precast panel. This provides lateral restraint for the support of the top of the precast however does not connect the PFC to the wall beam. The exact name has slipped my mind! but they are common.

Regards,

"Structural Engineering is the Art of moulding materials we do not wholly understand into shapes we cannot precisely analyse, so as to withstand forces we cannot really assess, in such a way that the community at large has no reason to suspect the extent of our ignorance." Dr. Dykes, 1976

civeng,

further to my last post they are called 'fixing clamp'.

regards,

"Structural Engineering is the Art of moulding materials we do not wholly understand into shapes we cannot precisely analyse, so as to withstand forces we cannot really assess, in such a way that the community at large has no reason to suspect the extent of our ignorance." Dr. Dykes, 1976

Thanks aaronPTeng,

I think I know what your refering to.

A ferrule in the panel with a clamp plate bolted to the ferrule and welded to the flange of the channel.
Still fixes the panel to the eaves channel and also the eaves channel must still be fixed to the roof beam to achieve the roof bracing which is where the moment transfer occurs.

Either way because the eaves connects to the roof beam there is always some moment transfer so the aim is to minimize it.

Thanks again.

aaronPTeng,
Clamping could take the normal wind force, but I would be dubious of that type connection for taking the strut force into the shear walls. I think there would have to be at least intermittent positive connections to transfer the in plane force.

Civeng,

I miss typed one part sorry, I do fix the wall beam to the roof beam, However Is there a need to weld the clamp plate to the flange of the PFC?

Regards,

"Structural Engineering is the Art of moulding materials we do not wholly understand into shapes we cannot precisely analyse, so as to withstand forces we cannot really assess, in such a way that the community at large has no reason to suspect the extent of our ignorance." Dr. Dykes, 1976

aaron

Yes it is necessary because

1. Of what Hokie said about transfering the strut load to the shear wall if you go with this type of connection.

2. Over time the nut on the ferrule may loosen (even slightly) and the clamp drop off.

civeng80,

1. This depends on the design, An I am guessing in your case for your situation you need this.
2. I dont think this is really the case As the base of the clamp is fixed hard up against the precast. I don't see how it can loosen. either way just a suggestion I have done plenty of connections this way and it is fairly typical, However each to their own

enjoy,

"Structural Engineering is the Art of moulding materials we do not wholly understand into shapes we cannot precisely analyse, so as to withstand forces we cannot really assess, in such a way that the community at large has no reason to suspect the extent of our ignorance." Dr. Dykes, 1976

These structures stay up for 50 years or more.

Many things happen in 50 years.

I always like to weld them (tack welded) then forget them.

Cheers

I don't like the cast in plate detail, N20 bar cranked into a 150mm panel for what purpose? N20 sounds large and the crank is going to be fun. I would suggest N16's would be better, and probably N12's would cut the mustard.

maybe something more like this??? http://www.reid.com.au/Reid_Services/Reid_products...

http://www.nceng.com.au/
"Programming today is a race between software engineers striving to build bigger and better idiot-proof programs, and the Universe trying to produce bigger and better idiots. So far, the Universe is winning."

I'm probably a bit late to the conversation, but I can tell you that I've seen a lot of this detail in New Zealand, along with a few failures. This detail failed in a fire scenario, costing the owner the ability to reconstruct, as well as plenty of failures under EQ load. Both due to the moment, but this is very rare indeed.

Frankly in Aussie I understand you have much less significant EQ loads, and I can see why these are common and no causing issues. I agree that there will be a moment induced, but I doubt it would be in any way significant. Let's not forget that what we are looking at in 2D is separated by a meter in real life... Small rotations and deflections will eat/dissipate a great deal of load in this case. P-Delta is not always our enemy.

FYI: Where I need to be able to predict behaviour accurately (ie: In NZ Capacity Design), I use the shelf angle for construction (only) and introduce a 15mm X 50mm X 200 or 300mm (depending on length required to extend past pair of or 4-bolt set) plate and guarantee pin behaviour.

That's likely over the top for Aussie, with your Cyclonic loading needing more "hold onto the weight" design and less "flex without breaking" seismic design.

Thats an interesting comment.
Can you elaborate how the connection caused failure in a fire due to the moment ?
Also could you place a drawing of the connection you use in NZ ?

Very interested in your comments.

So the failure which occured due to fire was in a large warehouse. Piles of stored wares were involved, and the heat caused the rafters to sag and twist. Once the beams were effectively counter-cambered, they were unable to keep from continuing to rotate, and the top bolts in the affected (discussed & shown above) connection blew out. The lower bolts never failed, and the total rotation at the knee was never beyond the capacity of a true pin connection. While I cannot know a true pin condition would have saved the day, it would have stood a much better chance, and would have likely saved the owner from replacing panels and rafters both.

I'm attaching a sketch of the detail I've used in NZ; I'm on parental leave at the moment and don't have a scanner at home, even if I did have a drawing of the detail. Note that this also comes in very handy where you are close to the limit on the shear value of the connection. You can't neglect the induced moment in the panel, but you can limit the design values for the bolts to pure shear.

<a href="http://tinypic.com?ref=1zd8y7d" target="_blank"><img src="http://i40.tinypic.com/1zd8y7d.jpg" border="0" alt="Image and video hosting by TinyPic"></a>

Thanks for that CEL.

So you use a shelf angle for erection only (then remove) and connect to what seems to be some steel packers to the panel ?

But what about the eaves channel which is the main culprit in inducing the moment ?

If the contractor wants the shelf angle to make their job a little easier, yes, it is for construction only. Most of the time I've found the panels are propped and the rafter is crane lifted, so they don't bother with the shelf angle (though I have seen it used, and required that it was removed afterwards).

For the eaves channel, the pin details means this issue is also solved. They are in line, or nearly so, therefore the moment is now zero and/or negligible.

Note that where required structurally, I have used back to back Diamond CFS channels powder fastened to the panel. Cheaper and quicker to build. If the loads are too high for powder fastening, I try to change the load path or rearrage to minimise load. Never been a big fan of too many post-fix anchors or chemsets. That many embeds is just going to lead to tears/site work/money.

Make sense?

Thanks again.

This topic interests me alot.

Ive been looking for a better more physical pin type connection for a long time.

The one you presented sounds good, but is it possible to see it in a bit more detail ?

Questions like is the butt plate to flange welded to shim on panel?

How do you pack up exactly to the beam butt plate etc.

Your comments greatly interest me.

The shim plate is welded onto the end plate, and then the gap (15mm to 25mm typically) is left to be dry packed.

I've also extended the end plate up to be able to add a final purlin where the panels are not kept tall (ie: no fire condition or other architectural requirement), thus having it do double duty. Otherwise you often see an akward detail with an extended cleat to hold a final purlin out and away from the knee. Looks like garbage and is even tougher to fabricate and install. Terrible detail, frankly!

To be honest I dont like this detail and I think it would be rejected by Aussie Engineers.
I haven't seen one detail that fits the bill yet.

Well, keep looking and keep working towards figuring out something you do like... Each detail should be thought through in great *cough* detail *cough*.

I love introducing the rocker (what you called a steel shim) because it lets me control the behaviour and I know that I won't get unexpected prying forces.

Let us all know what you choose to do in the end, but I really can't see why anyone would reject a detail that introduces a true pin condition... Would it be rejected on the basis of undue complication? Certainly can't be rejected for any reason of safety or predictability of behaviour.

Please don't be upset by my comments, maybe I didn't express myself well.

Its probably a good detail when its up and functioning.

Construction of of it may also be OK once you get a system going.

Are the bolts from panel to beam cast in ferrules or Trubolts or chemsets and what happens if you need 4 bolts for shear?

Hi CivEng80,

I'm not at all bothered by the comments, if anything having another engineer criticise your work makes you think again and build your skills!

The anchors are embeds (cast in elephant foot ferrules), and in the cases where I've needed four bolts, I've placed two to each side of the flange. Then you need to ensure you get the load spread, so I use a diamond shape end plate welded onto the end of the rafter to spread the load, rather than trying to check the rocker plate as a beam. The most I've used was eight ferrules in two rows, which then get checked for moment as a group, but permitted the rafter to see a true pin and minimised the design loading required at the connection.

See the attached sketch and calcs. This is how I have done tilt-panel and other concrete connections that support steel beams, and how I have analyzed them in considering the moment from the eccentricity of the connection (whether a seated angle, single shear plate, etc.)

At least in Florida, USA, welded headed studs have been one of the most popular forms of these types of embed plates. To deal with the reduced values of headed studs and other connections in concrete that are the result of ACI Appendix D (complex code that greatly reduced the capacity of concrete anchors), some engineers now use a shear lug concept, which I prefer for higher loads. This gets you out of Appendix D (for the shear lug), and then you can design using standard shear and bearing methodology. I may add additional horizontal or vertical rebar for crack control depending on the loads and type of concrete member, and have used additional rebar and the strut and tie method to reinforce the concrete under the shear lug to prevent a diagonal crack and shear failure.

Connections using epoxy or expansion bolts are useful in existing concrete, similar to the connection shown in the lower right corner of the sketch. The manufacturers now have some pretty good design programs in the USA at least, which can make design a lot easier and faster (multiple iterations). In any case, remember you may have shear and tension in the upper fasteners simultaneously.

Back to your connection, I may consider turning the TOP hooked bars horizontally, and then providing supplemental vertical bars to reinforce the possible shear failure of the upper rebar in tension. This would also give you a couple of additional vertical bars in the panel in the area of your connection where you have the highest stresses. Though an additional horizontal bar I indicated in the sketch with your detail would do about the same thing.

Hope this helps. Like to hear what other engineers are doing out there with these types of connections. They become very problematic with thinner panels and especially near panel edges, door openings, etc.

• http://files.engineering.com/getfile.aspx?folder=ffa983bd-da4d-4c65-8521-1b

A2mfk: Great post, but bear in mind that the added horizontal bar has nearly no effect on the SLS state and the initial (read: apparent Failure - initial cracking and the upset call from the client) and only very modest gains on the ULS. There is an excellent report on this from the University of Canterbury. I can post photos of my copy if anyone's keen.

keen

http://www.nceng.com.au/
"Programming today is a race between software engineers striving to build bigger and better idiot-proof programs, and the Universe trying to produce bigger and better idiots. So far, the Universe is winning."

Here you go rowingengineer... Apologies for the watermark; Like I said above, I don't have access to a scanner atm...

• http://files.engineering.com/getfile.aspx?folder=c2db8ed8-9acd-4495-9fb1-20

Back to the original question.

Been looking at a Bunnings building which is huge. It has the same connection as mine shown above but no bolts on the shelf angle that I could see at least. So the joint is completely free to rotate. Anyone involved in the design of these buildings here ?

Again any comments would be appreciated as the checking engineer is giving me a hard time on this.

The connection of the upper channel girt to wall panel is not shown.
If you can use the seat angle to transfer rafter vertical load to the panel, as well as panel lateral
loads to the rafter, you can eliminate the channel girt. This would be much closer to a "pinned" connection at the
top of the panel and avoid bending moment in the panel.
You will probably need to modify the seat angle details or sizes to transfer both lateral and vertical loads.

civeng80,
What shelf angle? I didn't see a shelf angle in your detail.

hokkie my apology,

That should be "angle seat".

So what does the "checking engineer" recommend? No good to be critical without participating in solving the problem.

Recomends portalising roof beams with panels. Different system altogether.

Yes, different system. More robust, though.

but more of a fire risk

http://www.nceng.com.au/
"Programming today is a race between software engineers striving to build bigger and better idiot-proof programs, and the Universe trying to produce bigger and better idiots. So far, the Universe is winning."

make that more of a risk in a fire

http://www.nceng.com.au/
"Programming today is a race between software engineers striving to build bigger and better idiot-proof programs, and the Universe trying to produce bigger and better idiots. So far, the Universe is winning."

Let me elaborate.

Not a steel portal frame with steel columns and beams.

Use the panel as a beam column and connect the roof beam with moment connection to panel.

It sounds OK in theory but I don't have enough literature on this type of construction to be comfortable with it

Definitely like steel portal frame though.

Panels as beam-columns with a Rafter's moment capacity is not going to work with an economical thickness of panel. You're going to wind up face casting a second lift to actually produce a column beneath the rafter. This is sounding wasteful and unnecessary.

I have seen this a couple of times, but only with very large (read: tall) panels indeed....

I don't like the sounds of it. a thin panel taking moment loads from a rafter sounds like hard work.

I would look to adjust your connection, I have made a few suggestions, if this is heavily loaded and the vertical angle is transferring to much I may look at the vertical angle and replace with a horizontal angle that sits on two angles/castin plates.

http://www.nceng.com.au/
"Programming today is a race between software engineers striving to build bigger and better idiot-proof programs, and the Universe trying to produce bigger and better idiots. So far, the Universe is winning."

• http://files.engineering.com/getfile.aspx?folder=1bbaba86-18c3-477d-8766-a5

I thought your tormenter was worried about distress in the panel. Now he want you to do a moment connection to it? Makes no sense.

RE, why do you think the steel portals present more of a risk in a fire than the concrete bearing wall system?

Probably getting at the fact that the concrete cover protects from fire issues, but the steel portals would have to be competently protected.

Alright, I'll bite as well: RE - What Hokie said!

Steel portals by themselves do not present more of a problem but when you clad with concrete panel this can be a higher risk in my opinion (however this risk can be handled by proper detailing, Fire ties to columns would be one such detail).

Reason is that the panels and steel portal will naturally want to bend away from the fire in the first instance in both forms of construction; however when the steel starts to sage the idea is that the steel pulls the panels back into the building reducing the risk to structures or property outside the building.

For the concrete panel option you are normally detailing a pin type joint at the bottom or allowing for plastic failure of the panel for this pull-in to occur before the rafters pull off the wall, there is a clause in the BCA to handle this connection capacity requirement to assume this happens.

For the steel portal option you need the columns and panels to form this plastic hinge for them all to fall in. I don't see this as very easy thing to make happen compared to the single panel, however if you detail fire ties at the top of the columns to allow the panels to blow off the columns and form independently the plastic hinge, I think there is a chance.

http://www.nceng.com.au/
"Programming today is a race between software engineers striving to build bigger and better idiot-proof programs, and the Universe trying to produce bigger and better idiots. So far, the Universe is winning."

Hokkie,

I have a lot of respect for this engineer.

His idea is either beam pinned to the panel or fully fixed to the panel.

According to him a pinned connection is statically determinate which is good for analysis and design but not so good for limit state design as failure could be catastrophic if one element fails e.g. a roof bracing member.

So his theory is make the structure statically indeterminate e.g. fixed beam connection to panel and to base and use this redundancy as a backup safety measure and moment redistribution, so that the structure would not fail catastrophically under extreme loads including any unforeseen loads e.g. blast loads, earthquake or impact load from vehicles (they do occur in industrial buildings).

It sounds good in theory but in practice I don’t know as there is not enough literature on this method including the economic viability of construction.

I must admit that there are some pretty poor tilt up panel buildings with a lot of problems out there, of which owners and builders remain silent on.

But then there are some pretty good ones (or so it seems) like the Bunnings warehouses.
I choose to make the connection pinned and statically determinate structure with roof bracing to hold the building stable (in reality of course no structure is truly statically determinate) . I think that these type of buildings and the Industry that developed them have been around long enough now to seriously consider them viable alternatives to the steel portal frames. I know AS3600 does not say much about tilt up and if read to the nth degree require more rebar in panels than what’s actually put in by the industry. (Mu > 1.2*cracking moment)

Rowingengineer thanks very much for going to the trouble of detailing alternative connection.
Thanks all for some pretty good thoughts on this topic.

The concept of redundancy due to portal action is great for steel structures, and arguably to a lesser degree for concrete structures. I just don't see that it is applicable to tilt up concrete panels.

Agreed! This is someone trying to fit a theoretical model to a practical reality... I like to have the panels be robust, the connections be of the seismic yielding type to the foundations, and the steel to panel connections be well overdesigned.

I agree, good in theory but in practice, got my strong reservations.

Well remember: Review Engineer or no Review Engineer, this is your design. Never allow something to go out that you don't believe in...