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Helpful Member!  slickdeals (Structural) (OP)
7 Oct 10 18:00
Folks,
When you calculate the uplift load on a footing, the code requires the use 0.6D + W. Assume this gives a net uplift of 30 kips.

When you calculate your resistance to uplift in terms of the footing weight and the weight of a truncated soil pyramid (based on a 30 degree angle), do you use a 0.6 factor on the uplift resistance and compare it to the above calculated uplift of 30 kips?

It seems like double dipping (very conservative) to use a 0.6 factor on the resistance side also. Thoughts?

dcarr82775 (Structural)
7 Oct 10 18:07
The 0.6 applies to DL and soil load resisting uplift.  There is no double dip so to speak.

Example:

- Uplift due to Wind = 50 kips
- assume resisting load of Dead+soil = 100 kips

100*.06 = 60 kips > 50 kips so ok.
 
Lion06 (Structural)
7 Oct 10 18:45
Agree with dcarr, the bottom of the footing is subject to rhe combination  0.6D + W and you can't have a net uplift.  I agree it's conservative, but it's at least partially to account for the F.S. of 1.5 that is no longer required.
msquared48 (Structural)
7 Oct 10 19:10
Used to be .9D + W with a FS of 1.5.  The relative degree of conservatism depends on the relationship between the allowed dead lload and the wind uplift.

Mike McCann
MMC Engineering
Motto:  KISS
Motivation:  Don't ask

frv (Structural)
7 Oct 10 20:50
I think Mike has it right..   

I've seen some engineers use 0.6DL+1WL and further use a 1.5 factor of safety..  That would be double dipping.
Helpful Member!  BAretired (Structural)
7 Oct 10 22:30
Yes, but they are double dipping on the right side...conservatism.

BA

slickdeals (Structural) (OP)
8 Oct 10 3:20
Actually my question was more regarding using the 0.6D factor on the weight of the footing and soil.

I was thinking along the lines of calculating the uplift on the superstructure above the footing based on 0.6 D and comparing it to the resistance based on 1.0 * (weight of footing and soil) and providing a F.o.S of 1.1 or so.

But I think the consensus is that I may be violating the letter (or spirit?) of the code.

ToadJones (Structural)
8 Oct 10 8:10
I see what you are saying Slick, and I have done it in the past, but usually when doing so I still had a pretty big F.O.S.
I never saw a reason to multiply the actually footing weight x 0.6.
The soil maybe, as it could be excavated. My biggest concern would be the dead loads of and on the superstructure.  
SteelPE (Structural)
8 Oct 10 8:54
dcarr82775 (Structural)
8 Oct 10 11:24
The new is the same as the old : (1/0.6)*0.9 = 1.5

They just re-wrote 0.9D with a 1.5 factor of safety into the standard load combination equation form.

There has always been a 1.5 factor against overturning conservative as it may be it has always been there and you need to follow the code.
SteelPE (Structural)
8 Oct 10 13:28
msquared48 (Structural)
8 Oct 10 19:46
... Or just run the numbers and prove it to yourself.

Mike McCann
MMC Engineering
Motto:  KISS
Motivation:  Don't ask

kikflip (Structural)
9 Oct 10 3:34
I would apply the 0.6 factor to all dead loads which have a stabilizing action on the structure.
Helpful Member!  ron9876 (Structural)
9 Oct 10 9:44
The code says 0.6D. It doesn't say 0.6 x the dead load that you would like to use. If you calculate the gross uplift and the total dead load you need a safety factor of 1/0.6=1.67.

I think it is an overkill. In hurricane areas with one story light framed structures the size of interior footings is not reasonable. When hurricane Andrew came thru I don't remember anyone reporting footings failing in uplift and I don't think many of the older buildings considered uplift on footings at all.

I think it is really crazy when you are designing something like a big shearwall pilecap. The design of the concrete and reinforcing steel is done with load factors so they have their safety factors. The piles have a safety factor from the geotechnical side. Then you add the 1/0.6 safety factor for the pile loads and you have safety factor x safety factor. There should be some type of allowance for this condition.
abusementpark (Structural)
9 Oct 10 14:25
To kinda pile on to ron9876's post, I think it becomes even more overly conservative when the geotechnical engineer wants you to take the buoyant weight of the footing and soil in consideration for uplift resistance.  So then to truly meet to the code, you have to multiply the bouyant weight of the footing and soil by the 0.6 factor.  In hurricane country, you can end up with some unreasonably sized footings.

Another monkey in the wrench, is whether or not you consider adhesion on the sides of the footing for uplift resistance.  In my area, some geotechnical firms give you an allowable adhesion (usually 500 psf) to consider acting on the sides of the footing for uplift resistance.  This can help tremendously and then using the 0.6* buoyant weight of footing and soil becomes less significant as you can usually get a significant contribution from the adhesion.

However, as I said above, for some reason, only a few geotech firms allow you to consider that, which has always been peculiar to me and could be a whole separate thread.  I've often found it too conservative to use the 0.6 factor, the buoyant weight, and no consideration of adhesion, which unfortunately would be the "by the book" way one would have to design a footing for uplift in certain instances.  
Helpful Member!  structuresguy (Structural)
11 Oct 10 10:03
personally, I always have a problem using the weight of the soil to help hold down the building, adhesion or no.  The reason is there is no guarantee what condition the soils will be in when they are most needed.  In fact, in hurricane country anyway, it is most likely that the soils will be in their worst condition when you need them.  I mean that hurricanes typically happen during the wet season.  So there is highly probability that the soils could be saturated during a hurricane.  The saturated soils will tend to "flow" off the top of the footing as it lifts up, due to the saturated condition, more easily then in dry condition.  This could also happen to saturated clays, particularly those subject to liquifaction.

So I generally don't use the weight of the soils, at least when I think it is reasonable to assume saturated soils during peak uplift event.   I prefer to rely on something "concrete", so to speak.  Something I can be 100% sure that it will do what I need it to do.  Besides, concrete in the ground is pretty cheap insurance.
ron9876 (Structural)
11 Oct 10 12:35
Boy you must get some really big footings. For me when I have a 8'x8'x4' footing to hold down an interior column in a retail space I can't see doing anything to add to that.
slickdeals (Structural) (OP)
11 Oct 10 14:14
@ron9876
I am in the same boat as you are. I have a 7.5'x7.5'x3' footing 3' below grade for an interior column in a retail building as well.

msquared48 (Structural)
11 Oct 10 15:05
Interesting.  You must be in a hurricane region with high uplift.

I am in a region where we frequently have to design failure panels for structures in flood areas to allow flood waters to pass through without failing the structure.

As a point of conversation, has nothing similar been either tried or proposed in hurricane regions to limit the uplift seen by the footings?  

Mike McCann
MMC Engineering
Motto:  KISS
Motivation:  Don't ask

structuresguy (Structural)
11 Oct 10 16:31
@ron9876:  Yes, occasionally I get some good size footings.  But since they are only 16" down usually, the weight of the soil above them is not that significant anyway.  Not like they were 4 feet down, with piers for the columns.  So adding an extra 8-12 inches of concrete gets me the weight of the soil above the footing.  But I also don't do a lot of large 1-story warehouse buildings, where I would have large column bays and little dead load to resist uplift.  

Just as a sanity check, are you calculating net uplift using MWFRS or C&C wind pressures?  We always use MWFRS for uplift.  But I have heard some people using C&C, which would result in much larger uplifts.
slickdeals (Structural) (OP)
11 Oct 10 16:44
I agree, we calculate the wind loads for structural uplift using MWFRS. The wind pressures drop drastically when you are >2h measured from the leading edge.

ash060 (Structural)
11 Oct 10 17:26
I asked one of the code writers for ASCE 7 this question, about very larger uplifts on the footing and if that was the intent of the code he said it was not, and he gave some guidance on alternatives.

One thing is that for alot of these one story buildings typically the columns are steel.  The anchors rods are designed for the uplift from the column, but not to engage the entire weight of the footing.  So the largest the footing should be is the maximum the anchor rods can pick-up.  Adding more concrete than that is just a waste because the column will pull-out of the footing before all the weight of the footing is engaged.  

For example the net uplift is 25 kips and the weight of the footing and soil on it is 40 k.  What is the point of that if the anchor rods are designed for 25 k, any more foundation than that is waste.  The 40 k will never happen.

So I have started designing for the maximum load that the structure can impose on the foundation, and not some number that the foundation can never see.

If Dorothy's house starts flying by my window it is time to start make our foundations larger, but until than maybe some engineering judgement is all that is called for.
slickdeals (Structural) (OP)
11 Oct 10 18:10
@ash:
And the hawk-eyed code officials in South Florida are cool with it? smile

structuresguy (Structural)
12 Oct 10 9:41
@ash:  I don't understand your logic.  If your net uplift is calculated by the appropriate Code mandated load combinations (say 0.6D + W), then your anchor rods will need to have a HIGHER capacity than just the actual weight of the foundations.

For example, if the gross uplift is 50 kip, and the superstructure dead load is 10 kip, the reaction at the foundation would be R=(0.6*10 - 50) = -44 kip.  So the foundation would need to weigh (0.6*Dfdn) = 44 kip => Dfdn = 73.3 kip.  So the anchor rods would need to be able to transfer all of this 73 kip load, since you are relying on all of it being there to counterbalance the uplift and satisfy the Code mandated load combination.  So the anchor rods should actually be designed for MORE than the weight of the foundation, not less.
ron9876 (Structural)
12 Oct 10 12:39
@ash: don't forget that the anchor bolt design has material safety factors included so they probably could pick up the full weight. Not saying that I agree with the requirements but I don't see how you could argue that you met either the intent or the letter of the code if you don't use 0.6D. I think if you were ever challenged you wouldn't be able to defend the position.
ash060 (Structural)
12 Oct 10 16:11
@structuresguy: Anchor rods are needed to keep the column down, not pick the footing up

@ron9876: I met the intent, but not the letter.  The footing is not going to uplift.  If it is a failure that has never happened does it apply?

@slick: They have never asked

I use 0.6D for all the superstructure.  The main thing I am trying to point out is that sometimes the code should not be defaulted to over engineering judgement.  Anyone can follow the code, but not everyone is an engineer.  
cvg (Civil/Environmental)
12 Oct 10 16:43
you guys are amazing! you mean that you don't always have to meet the minimum requirements of the code because you are an engineer? Than who was the code written for, all those non-engineer building designers?

I am a little hesitant to travel to south florida now. I expect that big hurricane that has never happened yet, to hit about the time I step foot off the plane...
kikflip (Structural)
12 Oct 10 17:40
8x8x4' seems a little excessice for an interior column in retail space, even for hurricane zones. What I do is tie the slab into the footing and provide top steel in the slab so it can cantilever from the footing. It is also a good idea to match the size of the anchors with the mass of the footing to ensure continuity of the system.
a2mfk (Structural)
12 Oct 10 18:52
I have spent my whole career as a structural engineer in Florida, mostly with low rise structures where this whole footing uplift deal is a big issue in design. I am relatively conservative as a structural engineer, but I see the two sides of this coin.

1. CODE:
I have argued there should be a small note added into the code to allow the use, at the discretion of the engineer of record, to use up to 100% (ok, maybe 90%) of the weight of an engineered slab and/or foundation.  

But right now, if we are talking about what "code" says, I think they all say 0.6D or similar. And we all know how perfectly written codes are (tongue firmly implanted in cheek). Its been my understanding that the 0.6 dead factor is because the majority of dead loads have unpredictable weights that may change throughout the lifespan of a structure. A foundation of course does not fit this description, and thus in my opinion, it does not fit the intent of the 0.6 reduction. I do not believe you would violate the spirit of the code in this sense, for whatever that is worth I am not sure :)

2. PRACTICE
I have been fortunate to perform forensic investigations after Hurricanes Charlie, Jeanne, Francis, Katrina and several tornadoes. The only footing failure from uplift I saw was arguably a scour/storm surge failure, but a front porch post embeded in a small cube of concrete was probably about 50 feet from where it was built (on the bay in Biloxi). I have seen lots of roof failures, and structures ripped from the slab and foundation. Even a partial roof cladding failure will drastically reduce the uplift load getting down to a footing.

I think we are all not being realistic if we think that wind pressure in a one story building will pull a footing out of the ground by punching a whole in the slab. So the building envelope would be perfectly intact- roof, windows, doors, etc. which would allow the wind pressure to continue to keep increasing? And no element along the load path would fail prior to the foundation being literally pulled out of the ground? If this happens, we have bigger issues to deal with, like the survival of the species.

I guess that would be my argument for the code change to allow 0.9D for foundations and slabs....



 
BAretired (Structural)
12 Oct 10 19:41
If you change 0.6D to 0.9D for ASD, you would have to change 0.9D to 1.35D for LRFD.  I don't know how that can be justified.

BA

ron9876 (Structural)
13 Oct 10 13:52
Andrew was in excess of the design loads that were in place at the time (and now). I was with a firm that was involved with the review of the damage and there were no discussion of footings that failed in uplift. There are many older buildings down here that didn't consider uplift on footings. 3'x3'x1' footings didn't fail in uplift.

@ash the lawyers would eat you up. You wouldn't have a defense. We don't have to agree with the code but when we put our seal on a set of drawings it means that to the best of our knowedge the design complies with the code. We all know that the number of handicap parking spaces required at the mall are excessive because we have never seen them all filled ever. I don't see where there is any difference here.
slickdeals (Structural) (OP)
13 Oct 10 13:57

Quote:

We all know that the number of handicap parking spaces required at the mall are excessive because we have never seen them all filled ever. I don't see where there is any difference here.
I see a big difference because that is probably not a life safety issue.

However, I agree with you that if you certify a set of drawings as meeting the requirements of a building code, then you cannot choose to exercise engineering judgment. Probably there is no defense for that.

On the contrary, every building that is built is not built in 100% accordance with the code.

BAretired (Structural)
13 Oct 10 14:36
The following comment was made earlier:

Quote:

The piles have a safety factor from the geotechnical side. Then you add the 1/0.6 safety factor for the pile loads and you have safety factor x safety factor. There should be some type of allowance for this condition.

I don't think there is any doubling of safety factors here.  A friction pile will have a skin friction value in pullout which is determined by the geotechnical engineer.  Total resistance to pullout would be 0.6*Dp + Skin Friction where Dp is the dead weight of the pile.  The skin friction value is not modified by the 0.6 factor.

BA

ron9876 (Structural)
13 Oct 10 14:55
@slickdeals yes but they had better be closer than a 67% overstress. There would be no defending that position.

BAretired my understanding is that the geotechs calculate the ultimate capacity of a pile and use a safety factor of 2.0 to determine the allowable capacity. That's what I meant. So you have the 2.0 safety factor and the 1.67 safety factor.
BAretired (Structural)
13 Oct 10 16:09
ron9876,

You have the 2.0 safety factor on pile friction and 1.67 safety factor on pile dead load, but you do not multiply  the two safety factors together.  What allowance would you make for this condition?

BA

ash060 (Structural)
13 Oct 10 16:25
Does anyone agree that the code is excessive in this case besides me?
BAretired (Structural)
13 Oct 10 16:30
I do not agree.  The code seems reasonable.

BA

ron9876 (Structural)
13 Oct 10 19:13
@ash: absolutely.

BAretired: you reduce the dead load of the building by 0.6 whci is same as a safety factor of 1.67 and then you resist it with a pile with a safety factor of 2.0.
abusementpark (Structural)
13 Oct 10 20:45
I agree with ron9876 here.  There are many instances where there is a doubling of safety factors.   
BAretired (Structural)
13 Oct 10 20:47
ron9876,

You should have been a politician!lol

Say the total uplift force is F↑ and the total dead load acting on the pile, including its own weight is D.  

Say that skin friction failure load of the pile is Fs,  so the geotechnical engineer gives you an allowable value of Fs/2.

Now, F↑ - 0.6D = Fs/2

or   F↑ = 0.5Fs + 0.6D

There is a safety factor of 2.0 for the skin friction and 1.67 for the dead load.  It is reasonable that a higher safety factor should be used for skin friction because there is a higher probability of going wrong than simply calculating the dead load.  But we are not multiplying the two safety factors together.  The resulting safety factor is somewhere between 1.67 and 2.0.

BA

BAretired (Structural)
13 Oct 10 20:52
abusementpark,

Where are these instances?  I know of none.

BA

abusementpark (Structural)
13 Oct 10 22:02
BAretired,

These instances occur, IMO, when you are applying the 0.6 factor to significant dead load that is GUARANTEED to be there, like in a foundation or concrete shear wall.

People argue that the 0.6D is still applicable because it implies a needed 1.67 factor of safety for overturning.  But this is only true for shallow foundations where you are relying on material weight for overturning resistance.  Consider a pile cap in which you are using piles for overturning resistance.  The uplift capacity of the piles has an additional factor of safety.  So, the way I see it, there are two factors of safety for overturning.

Personally, I think the code should be clarified to say that the 0.6 factor only applies to superstructure dead load.  The code can stipulate a 1.5 or 1.67 required factor for cases where overturning is being resisted primarily by dead weight.
BAretired (Structural)
13 Oct 10 23:05
abusementpark,

You  stated:

Quote:

I agree with ron9876 here.  There are many instances where there is a doubling of safety factors.  

Where is there a doubling of safety factors?  I find your argument totally unconvincing.  There is no doubling of safety factors and you are simply confusing young engineers who are trying to make some sense out of all of this.  Why don't you simply admit that you don't know what you are talking about?

BA

nutte (Structural)
14 Oct 10 10:41

Quote (BAretired):

I do not agree.  The code seems reasonable.

Not one to disagree with BA, I also think the 0.6D case is reasonable.  I don't believe the code gives you the option of increasing the factor (to say 0.9) just because you're sure of certain dead loads.  Skin friction resistance provided by a pile is not a dead load.  Again, I agree with BA's explanation above on this issue.

I found an abstract to an engineering journal article that deals with this topic, but I'm not able to access the paper itself.  Perhaps it would further the discussion, if someone here can get hold of it.

Counteracting Structural Loads: Treatment in ASCE Standard 7-05
J. Struct. Engrg. Volume 135, Issue 1, pp. 94-97 (January 2009)
Bruce R. Ellingwood, F.ASCE and Yue Li, A.M.ASCE
http://scitation.aip.org/getabs/servlet/GetabsServlet?prog=normal&id=JSENDH000135000001000094000001&idtype=cvips&;gifs=yes&ref=no

 
dcarr82775 (Structural)
14 Oct 10 10:56
@ Amusementpark:

I can not grasp the logic you and others are using.  Why is the DL of the superstructure any less known than the weight of the foundation?  The only thing that is unknown in the allowance for mechanical (or similar things).

- A W24x55 weights 55 pounds per foot, not 756, not 42.
- 1/2" gyp board weighs 2.2 psf, not 4, not 1.
- an 8" normal weight concrete slab weighs 100psf, not 75, not 150.

Assuming you have an 8" concrete slab type structure (apartment type construction).  Well over 90% of your DL is known to the same degree of certainty as the weight of your footing.  Yet for some reason you are perfectly willing to apply a 60% factor to this, but a 90% factor to the footing.  Why, how would you defend this in a court of law?  If you are carrying around an extra 10-20psf in your DL that is another story.

I have never seen a steel beam yield at 0.66Fy, so we should scrap that idea as well right?  I agree parts of the code are conservative, but I don't go willy nilly ignoring them because they make my life to difficult.
steellion (Structural)
14 Oct 10 11:31
All, it's hard to come to a consensus when we're discussing about 3 different things in one thread.

0.6D+1.0W is the uplift case. 1.0/0.6 = 1.67, that is your factor of safety.  This includes a little bit of "fudge" factor because the DL is not exactly known, and engineers tend to overestimate DL for gravity forces.  When this combination is used, it is NOT necessary to take on the additional 1.5 FS for uplift case, that's already covered.  The 0.6 has much more to do with providing an FS against uplift than knowledge of dead load.  If we really can't determine the DL within 40% accuracy, we've got much bigger problems on our hands.

I don't see why you would use this load combination at all for deep foundations.  The geotech will tell you that each pile has X allowable capacity of uplift.  Whether the uplift resistance comes from deadman weight or skin friction is irrelevant.  This already has FS included, so it's simply 1.0W/pile capacity = # of piles.
southard2 (Structural)
15 Oct 10 1:53
OK as one structural engineer in Florida who has forever been criticized for oversized foundations here is what I do in my practice.  I'm sure some of you will disagree with regard to code interpretation but this is what I do and I feel confident.

First my the .6DL for ASD design I think was intended to make sure the dead load wasn't overestimated with dealing with net uplift situations.  Usually for a gravity load situation in ASD the factor is 1.0.  For LRFD its 1.2.  So clearly the more confident we are on the accuracy of our loads the less of a amplification factor is required.  For live loads the amplifcation is much higher.

I think the 0.6 factor is extra low just because our base dead loads are usually overestimated for gravity load cases.  So by using a 0.6 factor we are sure not to be using too heavy a dead load when calculation net uplift.  Its not that all the sudden we are unsure of the dead load.

For single story buildings I actually calculate the net uplift at the base of the column using ASD's 0.6DL + WL. That is the acting force on the foundation.  On the resistance side I will then use the soil weight, concrete footing weight, etc... then make sure that the this resistance is 1.5 times greater than the net uplift from above.  My view is that the soil weight and the concrete weights are known and therefore don't have to have the 0.6 DL factor.  If I were to do this than I wouldn't think any factor of safety would be required.  

In addition I make sure that the concrete weight alone without any soil weight yields at least a factor of safety of 1.0.  That way if the soil conditions are soaked with water I still really don't have a problem.   Also if the foundation starts to lift out of the ground the water bouyancy issue goes to zero effect.   If the area of the column is greater than 800 square feet I'll sometimes use the MFWRS uplift loads since they are lower.  It just depends on my mood.

I'm sure some will disagree with this approach but it is substantial and I'm confident I'll never have a footing pull out of the ground.  I never considered cohesion cause I don't have time for that. What I do know is that I get accused of over designing my foundations on a regular basis and it gets really old explaining to someone why.  Every now and then you'll talk to a contractor whos seen overturned steel dumpsters from a hurricane and they get the idea.  I always try to paint the picture for people by having them imagine the building upside down with the columns hanging from the foundations.  Now hang about 10 cars on the columns, etc, etc...  You start putting the loads in terms of how many cars and they start getting the idea.  Still I see relative to mine plenty of light foundations out there that obviously have not considered uplift.

For the anchor bolts sometimes I'll use the proper LRFD factors but in most cases I'll just take the 0.6DL + WL and multiply by 1.6 and 2.0 which is almost always way conservative.  I see no reason why connections shouldn't be over engineered since they are cheap.  

I have to say though that I've never heard of a foundation pulling up out of the ground.  Not even on old buildings where certainly the engineers of old didn't even consider this.  My guess is that if one column starts to lift up all the sudden it carries a much larger areas of DL then just its normal tributary area.  Other columns might not even be in uplift at the same time cause we are talking about gusting winds.  So you need to have multiple columns foundation fail simultaneously.  The ASCE wind loads are probably way conservative with respect to the higher tribuatary area situations.  On top of that the cladding tends to probably fail first due to human error, higher localized loads, etc...

I've also seen those beat up metal building foundation engineers using some pretty low factor of safeties out there.  I caught one guy using 1.1 while including slab weights far the columns.   Be careful on those people.  I'm so sick of this net uplift issue that I've come close to no longer even taking on metal building foundation projects.  I probably would just start telling people I don't want them anymore because of liability issues only the economy is so slow with respect to new building construction right now.

 

John Southard, M.S., P.E.
http://www.pdhlibrary.com

AVak (Structural)
15 Oct 10 7:30
I am not sure how you can be sure that your foundation concrete is 100% there when it is cast into an excavation that often has an uneven bottom surface and sides.  It is no less variable than the building dead load itself.

As others have pointed out, the 0.6DL is intended as a replacement for the old 1.5 safety factor for uplift.  It results in a larger factor of safety of 1.67, but it cannot be ignored.

Adam Vakiener, P.E.
ron9876 (Structural)
15 Oct 10 13:04
BA let's say you have a shearwall pilecap that has tension on a pile. You can reduce the tension by using 0.6D. That is the first safety factor. Then you need to compare the net tension to the pile capacity whose ultimte capacity has been reduced by a safety factor. That is the second safety factor. They are additive not multiplied as I said above.

southard2 I have also had a similar problem on metal building foundations. I had an architect tell me that he knew an engineer that told him that he would use 1.0. I explained the code to him and told him that was the way it had to be. It worked that time.

 
slickdeals (Structural) (OP)
15 Oct 10 13:20

Quote:

BA let's say you have a shearwall pilecap that has tension on a pile. You can reduce the tension by using 0.6D
Shearwall pile cap with overturning moment will have more tension if you use 0.6D.

structuresguy (Structural)
15 Oct 10 16:16
Like others, I have never seen a foundation pull straight up out of the ground.  I have been working in Florida for 10 years now, and did quite a bit of forensic work during the hurricane year.  But one of the big reasons I think that we have not seen too many foundation failures isn't that they are over designed (well, any worse than a FS=1.5 anyway).  It is that the cladding systems of old, and even fairly recently too, fail much sooner.  This is why there has been such a big push in the recent past to ensure continuous tensile load paths from roof to foundation.  

I surveyed some heavily damaged buildings in Ft Lauderdale in 2004, metal buildings at the airport in particular.  The structure of the building was intact, but the roof panels were completely gone over most of the roof, and the wall panels were probably 50-75% gone.  The foundations were never given a chance to fail, or even come close to it.  Of course, everything inside was a complete write off.  It's just like that video showing the entire roof of some house coming off the walls completely intact.  

As cladding systems, and their connections to the structure, get better, and in-line with Code required load combinations, then the likelihood of foundation uplift failure will get much higher than in the past.  Based on this assumption, it is more important to ensure adequate safety factors in the foundation now, then in the past.
BAretired (Structural)
15 Oct 10 16:56
ron9876,

The safety factors are neither additive nor multiplied together.  They are separate safety factors for separate types of resistance.

slickdeals,

I'm not sure what your last sentence says.  Can you elaborate?

structuresguy,

I think you are correct in your assessment.

BA

slickdeals (Structural) (OP)
15 Oct 10 17:12
Say you have a moment M due to overturning in combination with dead loads. The moment gets resolved into tension and compression on the piles in a pile cap.

Using a 0.6D+W instead of D+W will produce more tension in the piles.

I could have misinterpreted Ron9876, forgive me if that's the case.

BAretired (Structural)
15 Oct 10 17:26
Okay, slick...I agree with that.

BA

abusementpark (Structural)
16 Oct 10 15:53

Quote:

Where is there a doubling of safety factors?  I find your argument totally unconvincing.  There is no doubling of safety factors and you are simply confusing young engineers who are trying to make some sense out of all of this.  Why don't you simply admit that you don't know what you are talking about?

I'm not sure why you feel the need to continually condescend people on this site. I'm not going to respond to it.  If there a specific statement of mine that you don't agree with, then I'd happy be to engage in civil discourse regarding the matter.   
abusementpark (Structural)
16 Oct 10 16:42

Quote:

I can not grasp the logic you and others are using.  Why is the DL of the superstructure any less known than the weight of the foundation?  The only thing that is unknown in the allowance for mechanical (or similar things).

As others have said, I think most engineers tend to take conservative estimates of the equivalent uniform load to be added to account for architectural ceilings, electrical, mechanical, etc.  I'm not sure how many sharpen their pencils to determine the true load added by your typical, light architectural coverings and finishes.  Obviously, if you know something significant is being placed (i.e. granite flooring), then it should accounted for specifically.

That being said, I think the largest aspect of uncertainty associated with the superstructure is what can happen when the occupancy changes from its original use.  A new tenant may decide that they want exposed framing members and they remove the entire architectural ceiling. Or a heavy floor covering may be removed and replaced with a much lighter floor covering.  So, we could crack our knuckles and determine an fairly accurate estimate of the true dead load for the structure right after it is constructed, but who knows what the future holds and when an extreme wind or seismic event will occur.  So, this is why I think the superstructure DL has more uncertainty then something you can pretty much guarantee is not going anywhere without extensive renovation and engineering, like all foundation elements or a major structural element like a concrete shear wall.

Quote:

All, it's hard to come to a consensus when we're discussing about 3 different things in one thread.

Exactly.  What exactly is the idea behind the 0.6 factor?  Is it to account for DL overestimation, DL uncertainty in the future, or both?  Or is it to provide a factor of safety for overturning when you are relying solely on the weight of foundation?

For light single story structures, I often do gravity design based on a DL,max and uplift design based on a DL,min, which is what I think the minimum weight of the roof structure could be.  Is it overly conservative that I multiply my DL,min by 0.6, since the uncertainty may have already been accounted for in the 0.6 factor?  Someone may say I don't need to take the "fat" out of my dead load for uplift calculations, since that is what the 0.6 factor is there for.
BAretired (Structural)
16 Oct 10 20:04

Quote:

What exactly is the idea behind the 0.6 factor?  Is it to account for DL overestimation, DL uncertainty in the future, or both?  Or is it to provide a factor of safety for overturning when you are relying solely on the weight of foundation?

It is all three of the above.  If you knew precisely the magnitude of the dead load and knew it would never change, would you change the factor to 1.0?  That would provide no safety factor against overturning.

BA

BAretired (Structural)
16 Oct 10 23:29
abusementpark,

This is what you stated on Oct. 13, 2010 at 20:45

Quote:

I agree with ron9876 here.  There are many instances where there is a doubling of safety factors.

Would you be so kind as to elucidate on where these many instances occur?  As I mentioned earlier, I know of none, but if there are some instances then I think they should be addressed.  Now is the time to speak up.

BA

ron9876 (Structural)
18 Oct 10 12:48
@slickdeals that is what I meant.

I have always thought that the need for a safety factor is the potential for the wind uplift to be larger that code minimums. Say one of those 155-160 mph storms makes it way on shore.  
abusementpark (Structural)
19 Oct 10 20:53

Quote:

Would you be so kind as to elucidate on where these many instances occur?  As I mentioned earlier, I know of none, but if there are some instances then I think they should be addressed.  Now is the time to speak up.

See Ron's post on October 15.  

Basically, anytime you have uplift being reduced due to the 0.6D factor on something that is guaranteed to be there (i.e. footing, pile cap, concrete shear wall, etc.) along with a geotechnical factor of safety on the uplift resistance of a foundation element (i.e. friction piles, drilled shafts, adhesion on footings).  For all other service loadings, we are comfortable relying on the geotechnical factor of safety alone.  Here we are factoring up the load by virtue of this 0.6D factor in addition to the geotechnical safety factor. That seems inconsistent to me.  If the geotechnical factor of safety alone is good enough in other scenarios, then why isn't it good enough here?    

Quote:

It is all three of the above.

Then that is the problem in my opinion. We are trying to use one number to cover a multitude of scenarios.  It seems inconsistent to me.  The differing levels of structural reliability depending on your situation could be over the place.
   
BAretired (Structural)
20 Oct 10 0:00
abusementpark,

If I have offended you, I sincerely apologize.  It was not my intention to do so.  

The Canadian code does not recognize the 0.6DL + W criteria because ASD (Allowable Strength Design) is no longer an acceptable standard in Canada.  The standard we use is LSD (Limit States Design) which corresponds to your LRFD.  It has been about thirty years since I used ASD and I may be a little rusty so please bear with me.

The equivalent load combination in LSD is (1.25D or 0.9D) +1.5L.  When dead load contributes to the load under consideration, it must be taken as 1.25D to provide a safety factor against collapse.  When dead load acts in opposition to the load being considered, it must be taken as 0.9D to account for the possibility that it may have been overestimated.  Live load must be taken as 1.5L when it contributes to the load and zero when it acts in opposition because live load is transitory.

Quote:

Basically, anytime you have uplift being reduced due to the 0.6D factor on something that is guaranteed to be there (i.e. footing, pile cap, concrete shear wall, etc.) along with a geotechnical factor of safety on the uplift resistance of a foundation element (i.e. friction piles, drilled shafts, adhesion on footings).

If there is a net uplift on a pile, there must be sufficient reinforcement between the column and pile to safely resist the uplift.  Using ASD, that would mean designing the reinforcement for an allowable stress of about 60% of it's yield.  Using LSD, we would take the factored uplift and design the steel for 90% of yield.  Either way, we would arrive at the same area of steel.

Now, column uplift has been delivered to the foundation. If the foundation weighs exactly as much as the uplift, then theoretically everything is balanced, but there is no safety factor against possible increase in uplift.  This is not in keeping with the principles of structural design.  A safety factor of 1.5 is needed.  This means that the foundation must weigh 0.67W where W is the maximum uplift force expected.  To allow for possible differences in unit weight of concrete, volume of concrete, etc. your code uses the figure of 0.6W.

The dead weight of the foundation contributes 0.6D to resisting uplift using ASD or 0.9D using LSD.

The remainder of the required uplift must be made up using skin friction of the pile or adhesion of the footing with an appropriate safety factor, as determined by the geotechnical engineer.

BA

Joepaterno (Structural)
15 Nov 10 16:29
I'm one of those younger structural engineers that is now confused because of this thread.

I believe I have an example that perhaps will spur some more interesting debate.  I have a stack, 150' tall, 32' in diameter, on a massive, rigid pile-cap foundation. The top of concrete is above grade, and no embedment effects are considered on the edges of the pile cap. I've been supplied with vendor loads for the stack as a whole. I've completed the design, and believe it to be adequate, but I think this provides a good example for the discussion.

Here are my design parameters:

Vendor Supplied Stack DL:             337 kip
Total  Foundation DL:                 1005 kip
Wind Load Overturning Moment:         7663 kip*ft
Bouyant Force on Bottom of Concrete:  380 kip

Using a STAAD model, I've identified that the tension in my piles is the governing failure mode.  Using the load combination 0.6x(DL stack) + 1.0x(DL Pile Cap) + 1.0xWL + 1.0xH (BOUYANT FORCE)

Note that I took the full load of the mat to resist the overturning.  I acheive a pile tension of 13.3 kips. My allowable pile tension, as reported in the geotech report is 13.5kips.  This allowable load already includes a 2.5 geotechnical capacity saftey factor.

Considering both a) the foundation DL is known with a high level of accuracy (our constuction tolerance make the design volume of concrete a minimum) and b) the geotechnical allowable capacity of the pile includes a factor of safety > 2.0, is this design acceptable?

I argue that this design is valid, in no case can the wind load from the stack be applied to the foundation with less than the full dead load of the mat being there. I used 0.6 for the DL of the stack, as it is conceivable that the stack liner could be taken out during maintenance. Additionally, the 1.5 factor of safety for foundation overturning the 0.6 factor is supposed to replace does not apply, as my foundation is pile supported.

Note that I have a tight site, with no chance of expanding the foundation plan area and spread out the pile group.  I can't see increasing an already 5'3" thick mat, costing the client tens of thousands of dollars, as the intent of the code, in using only 18% of the foundation weight to when calcuating the tension in the piles due to wind. (150 pcf * 0.6 - 62.4pcf = 27.6pcf, 27.6/150 = 18%)

Anyone agree / have material I can use to prove that the codes use of 0.6 for the DL of mat is absolutely necessary?
BAretired (Structural)
15 Nov 10 17:43
Joepaterno,

In my opinion:

DL = 337 + 1005 - 380 = 962k
0.6*DL = 577k

If N = number of piles, then each pile has an allowable resistance of (13.5 + 577/N) kips.   

BA

csd72 (Structural)
16 Nov 10 6:03
This may have been covered but:

Have you included the weight of the slab on ground in your resistance calculations? we used to use a distance of ten times the thickness of the slab in each direction.

Also for a very small added benefit, your internal pressure is also pushing down on the ground as well as up (but yes it is very small.

I have also seen people use a nominal amount of friction around the sides of the footing.
Lion06 (Structural)
16 Nov 10 6:21
Joe-

I think you still have to use 0.6 times your pile cap weight.  It's a DL and the code doesn't allow you to decide which DL is overblown and which is on the money.  That's the way I rerad it anyway.  When in doubt I always err on the conservative side - it helps me sleep better.
steellion (Structural)
16 Nov 10 9:00
Joe, I'd use 0.6 x Dead.  It's been debated on this thread, but I feel that the 0.6 factor has much less to do with the lack of knowledge of the true Dead Load as it does with providing an adequate safety factor.  Remember, while the Dead Load is well-known, the Wind Load and Buoyancy Force are highly variable.

The factor of safety for the piles are for individual piles, backfigured from the yield point.  I think you would still need the factor of safety against overturning that the 0.6D+1.0W+1.0H provides.
structuresguy (Structural)
16 Nov 10 9:23
Yep, I'm with Steellion and EIT, you have to include the dead load of the pile cap with the dead load of the stack, and multiply it all by 0.6 factor.  You don't get to decide which dead loads you want to apply this factor to.  Not when the code sl clearly mandates the load combination.  TO do otherwise opens you up to significant risk, should anything ever happen to your structure.  If you did not follow the code, you would surely lose at trial.
Joepaterno (Structural)
16 Nov 10 16:59
Thanks for your opinions, but I'm not sure I agree.

Using the logic that the 0.6 D was intended to replace the 1.5 geotechnical factor of safety for mat foundations for uplift, which is no longer used, why should the pile geotechnical safety factor for tension be used? (Keep reading, I know how that sounds)

If this was a mat foundation, I would simply use the 0.6D, and check against the tension with no factor of safety.  But it would be acceptable conversely in previous codes to use 1.0 (or 0.9 D) and account for the factor of safety at the end.  Similarly, the factor of safety my design counts on is embedded with the pile geotechnical allowable load. This allowable load has a 3.0 factor of safety. I don't think this allowable load should be checked against service level loads that have that inherent safety factor. It would be equivalent to taking a factor of safety to the applied load, and a factor of safety to the resistance. Total factor of safety would be 1 / 0.6 * 3 = 5, which seems very high.

Additionally, I've made the following conservative assumptions:

1) Water Table is actually at bottom of foundation, the full bouyant load on foundation is taken to account for extremely unlikely event of the water table being top of ground surface.

2) No cohesion taken on sides of 5'3" thick mat in cohesive soil.

Did I change anyone's mind?

 
Lion06 (Structural)
16 Nov 10 17:37
Not yet.  There's more than the geitech safety factor to consider.  Even if I were to agree with your approach, you still need to the 0.6 on the cap for the connection of the pile to cap, for the cap itself, and for the actual pile capacity.  The only thing your approah gets you out of is the required rock socket length (or whatever the transfer mechanism may be from pile to soil/rock).  Everything above that still needs the 0.6 factor on the cap self weight - even usin your approach.

If you have a footing with a large overturning moment (from wind) such that the resultant axial load is close to the edge of the footing (but still within the footprint of the footing) and using a 0.6 factor on the footing self weight results in the soil bearing pressure being too high would you still take the same stance and say you can use the full footing weight without the 0.6 factor since the soil bearing has a safety factor on it already?  I wouldn't and I don't think that's the intent of the code.
Joepaterno (Structural)
16 Nov 10 18:19
Also, in the California Building Code, the alternate allowable load combinations do permit the use of 0.9D per Section 1605.3.2
BAretired (Structural)
16 Nov 10 18:33

Quote:

Vendor Supplied Stack DL:             337 kip
Total  Foundation DL:                 1005 kip
Wind Load Overturning Moment:         7663 kip*ft
Buoyant Force on Bottom of Concrete:  380 kip

The dead load is weight of Stack + Foundation - Buoyant force:  i.e. DL = 337 + 1005 -380 = 962 kips.  So 0.6*DL = 577 kips.

The soil report provides an allowable tension of 13.5 kips based presumably on skin friction.  That cannot be changed by anyone but the soils engineers.

If there are n piles uniformly spaced around the periphery of the pile cap, each pile is capable of resisting an allowable tension of (13.5 + 577/n) kips.

The wind will stress some piles in tension, others in compression.  The greatest uplift will occur to one or two piles in line with the wind direction and the center of pad.  For those one or two piles, the calculated wind uplift should not exceed the value given above.

Does the foundation satisfy that requirement?

BA

csd72 (Structural)
17 Nov 10 11:45
load factors are for variability in loads, material factors are for variability in materials.

The minute we start to think that things will be built/behave exactly as we design them is the minute that we start to get failures.
BAretired (Structural)
18 Nov 10 19:16
Joepaterno says:

Quote:

If this was a mat foundation, I would simply use the 0.6D, and check against the tension with no factor of safety.  But it would be acceptable conversely in previous codes to use 1.0 (or 0.9 D) and account for the factor of safety at the end.  Similarly, the factor of safety my design counts on is embedded with the pile geotechnical allowable load. This allowable load has a 3.0 factor of safety. I don't think this allowable load should be checked against service level loads that have that inherent safety factor. It would be equivalent to taking a factor of safety to the applied load, and a factor of safety to the resistance. Total factor of safety would be 1 / 0.6 * 3 = 5, which seems very high.

This is faulty logic.  Resistance to the applied wind load is resisted by two separate entities, dead load and skin friction, each taking part of the applied load. For a given pile:

  0.6*DL + Allowable Friction = W(service load)

The allowable friction is determined by geotechnical considerations.  In this case, it is 13.5 kips/pile.  The factor of safety for soil friction may be expected to be substantially higher than the assessment of dead load because soil properties are highly variable.  It may be as much as 2.5 or 3.0 but it applies only to the term "Allowable Friction", not to DL.  Combining the two factors to form a total factor of safety of 5 is absurd.

BA

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