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capacity check - shear links
2

capacity check - shear links

capacity check - shear links

(OP)
Forgive me here but this is relevent to a British standard clause and I realise most here are US.

BS8110 (Concrete) states that all tension bars in a beam must be within 150mm of a vertical leg. It also gives other min dims. I don't know if you have similar clauses in US.

My question is this;

I am looking at an existing stucture for some increased loading, the check calls for links for the shear. The area is fine, however the link arrangment does not meet the detailing clause above. I am guessing that maybe in an older code version this was acceptable or the original design didn't need links so they put in the minimum area (albeit not to BS8110 detailing rules, but they could in theory be omitted totally).

To my mind the links are required to halt any shear crack propogating down to the main bars and reducing the capacity of the section be breaking the bond between the main steel and the concrete, Leading to failure. Hence the spacing only 150mm from the tension bar. Therefore in my assessment the non confoming detail will mean the beam in idadequate in shear due to the detail even though the area is above that required.

Any views?

RE: capacity check - shear links

2
A scheme maybe would help ... I think you refer to stirrups or ties as links, and the stated minimum separation for some case, but would love to ascertain if such is the case.

RE: capacity check - shear links

(OP)
sorry another difference in terminology, yes a link is what you would call a stirrup. Vertical or inclined bars to resist shear. I have attached a sketch. Basically the area of stirrup steel is ok, but the positioning does not meet code. I am trying to assertain the problems associated with this.

RE: capacity check - shear links

I can't help with the British Standard requirement.  The Australian Standard AS3600 specifies a maximum transverse spacing of stirrup legs to be the lesser of 600 mm and D, which is a less restrictive clause than yours.

RE: capacity check - shear links

(OP)
so if you were asseesing a structure and it had greater spacing would you take the links as uneffective and neglect all the links in your analysis?

RE: capacity check - shear links

No, I wouldn't neglect them.  I would try to research the impact the variation from the code provision has on the capacity of the member, just as you are doing.  I just can't tell you why your code has that provision.

RE: capacity check - shear links

(OP)
how does your code approach it if the detail is as described?  

RE: capacity check - shear links

by spanish code EHE 2008
stirrupus separation st

st ≤ 0,75 d (1+cotgα) ≤ 600 mm if Vrd≤Vu1/5

where α angle of the stirrup or shear bar with the axis of the member
Vrd design shear
Vu1 limit in shear of the section for compression of the web

st ≤ 0,60 d (1+cotgα) ≤ 450 mm if Vu1/5≤Vrd≤(2/3)Vu1

st ≤ 0,30 d (1+cotgα) ≤ 300 mm if Vrd>=2/3 Vu1

Shear is not a vertical section issue and I think I have read statements in a number of books where any shear reinforcement be thought to have shear containment effect on a length around where it is, that may be a wa of saying that an analogy of Mörsch truss scheme can adapt a bit to the actual position of a stirrup. Other thing is that following strictly the more recent codes one is more likely to get some reliable degree of assurance against unsightly cracks, starting for service level and beyond. But it is clear that typical cracks in shear maybe even more likely around 30 deg than 45 and so the crack will affect beyond d from the face, and any stirrup in such beyond d will have some effectiveness, specially for limit states.

By the way, in Spain the limits for separation of stirrups have been relaxed from 1998 to 2008. In 1998 were

st ≤ 0,80 d (1+cotgα) ≤ 300 mm if Vrd≤Vu1/5

st ≤ 0,60 d (1+cotgα) ≤ 300 mm if Vu1/5≤Vrd≤(2/3)Vu1

st ≤ 0,30 d (1+cotgα) ≤ 200 mm if Vrd>=2/3 Vu1

I really would agree more with this older statement than the new since trying to help to restrain a strut along some surface of the web, tighter spacings will be better.

Respect the BS code better one practicing there answer, I would have to make some reading.

RE: capacity check - shear links

Herewego,

Can you give a code clause that states that all tension bars in a beam must be within 150mm of a vertical leg.

I am not aware of this. I know that compression rebar in beams and columns needs to be within 150 from a link. This is to contain the compression rebar.

The detail that you post looks fine. I would include the links in your shear check calcs.

RE: capacity check - shear links

(OP)
BS 8110-1 clause 3.4.5.5  

RE: capacity check - shear links

BS8110

3.4.5.5 Spacing of links (see Table 3.7)
The spacing of links in the direction of the span should not exceed 0.75d. At right-angles to the span, the
horizontal spacing should be such that no longitudinal tension bar is more than 150 mm from a vertical leg;
this spacing should in any case not exceed d.


EC2 9.2.2 Shear Reinforcement

(8) The transverse spacing of the legs in a series of shear links should not exceed st,max:
Note: The value of st,max for use in a Country may be found in its National Annex. The recommended value is
given by Expression
st,max = 0,75d ≤ 600 mm


I am unsure how the values are derived but I would guess the code requirement is to ensure the shear can be adequately transferred to the links.  A particularly wide member, that requires links for shear strength, would otherwise have to all off the load to the perimeter links through varying stiffness material.

It seems that EC2 follows the Australian Standard, so you might not meet the current BS standard you might meet the standard after 2010 once 8110 is withdrawn and replaced with the Eurocode.

RE: capacity check - shear links

(OP)
nice try ussuri but I've already gone down that road. It doesn't meet the eurocode criteria or BS5400 (bridges). The beam is 1500mm wide. The links must have some affect but if I follow BD44 which is an assessment guide for bridge decks it says they are ineffective, but I take that as a conservative code cop out.  

RE: capacity check - shear links

having a look through some of design guides, one of which states the limits in EC2 come from test data.  The spacing has been chosen to ensure that a shear failure plane cannot be formed between two adjacent sets of shear reinforcement (designers guide to eurocode 2 by Hendy and Smith, p279).

If you dont meet the requirements as demonstrated by the EC2 testing, then although you have enough steel in certain locations, maybe you dont have it in the right places to distribute the load and prevent an 'internal' shear failure?  With some digging you might be able to find out what testing was done.

If this is based on reliable test data, it may be that the BS8110 approach was just based on the PPI principle (put plenty in) so is very conservative.  I dont have any 8110 references to hand.  Alternatively it might be a throw back to CP110.

Does your 1500mm wide beam only have perimeter links?

RE: capacity check - shear links

(OP)
yes it does only have perimeter links, so 2 legs on each edge of the section.

I don't belive it even meets the requirement of CP110 which was less onerous. I am going to look into the handbook to 8110 when I can track one down.

I believe all shear equations come from test data rather than any exact theoretical principle and are very empirical.

One proposal given to me was to calculate the Vc (concrete shear capacity for US members) taking only into account the tension steel within the 150mm limit of a link, in my case only 4 bars. Then combine that capacity with the links. Not to sure of this approach?  

RE: capacity check - shear links

I assume you mean the shear capacity is dictated by two thin sections (150mm+50mm?) either side of a 1500mm wide beam.   How would you demonstrate any load carried on the 'middle bit' is transferred to the perimeter where the shear strength is?  Maybe unless your arrangement was such you could argue that structurally you had two beams and the concrete in between was just fill?  I dunno.

RE: capacity check - shear links

(OP)
No I wasn't convinced either to be honest. seemed like a fudge at best. The problem I've got is due to what is going on the structure will definatly be loaded as I am analysing for. Its not a theoretical live load like is applied to normal floors and beams.  

RE: capacity check - shear links

So:

1) your applied load and associated concrete shear stress means your require links.
2) your applied load is fixed
3) You have sufficient area of steel for shear, its just all out at the edge of 1.5m wide beam, with no transverse steel across the beam section
4) detailing as existing meets no code requirements

Problem.  Demonstrate the section as existing has adequate shear capacity even though outwith the old, current and future design codes (and some foreign ones as well)???
  
Thoughts:

Can you strengthen the member?

Your load is fixed, can you reassess the load path or provide a different one by stiffening something else up?

Assuming no links and limiting shear stress to 0.5Vc is the capacity well under your applied (I realise this is for minor members)?  

Ultimately you have a reinforcement arrangment which is outside the scope of the codes.  Now as you point out the capacities and formulae in the codes are based on physical tests based on a set of parameters (cage arrangements) which you dont meet.

Do some tests yourself??? Not realistic.  

Demonstrate capacity theoretically??  I'm not sure how.

Talk to some of the people who did the tests? wrote the design guides?  or maybe try the concrete centre and the like?

Am i waffling? probably.

 

RE: capacity check - shear links

Anyway a width of 1.5 m without inner shear reinforcement seems too much. So your intention of looking at it as not acceptable sounds right. The analogy of Mörsch is better understood applied regularly on the mass of the beam. Otherwise you will need to ensure that through transverse strut and tie action (because if I would put deep-beam action reinforcement along the web -here inner reinforcement- is normally also understood to be required) where the tie is the inferior horizontal leg of the stirrup the tributary load is properly transferred to the stirrups proper etc. Also, it might be a strut and tie model at the sloped 45 or 30 deg of the compression strut, or whatever some truss analyses may say is adequate. It might work and still not meet the code.  

RE: capacity check - shear links

How deep is your 1500 wide beam?  If it is a wide, flat beam, you should be able to use the one way slab provisions rather than the beam provisions.  We often use "band beams" which require little shear reinforcement, and the ligs provided are only for support of the top steel.

RE: capacity check - shear links

If the beam is satisfactory for current loading with only one stirrup, then the shear stress must be very low, almost to the point where you do not require stirrups.

But if you want to increase the loading on the beam, you may want to consider drilling through the beam and inserting through bolts.  Is the flexural steel adequate to carry the additional loads?

BA

RE: capacity check - shear links

(OP)
I will try and answer all points;

BAretired - the existing design loads require minumum links only but do not meet code in the arrangment.

The beam is 1000mm deep.

This is definatly a beam rather than a slab I would say. there are no alternative load distrubution paths. It is a beam sat on piles.

There is no other load path for what the client wants to do.

The new load requires design links and not minumum only.

there is a difficulty in doing any strengthening works due to ownership/liability issues.

think I'm stuck!!!

RE: capacity check - shear links

Following with the strut and tie scheme to the stirrups at the sides ... it is not to forget that apart from checking the tie the compressive struts are also to be checked. Furthermore, the aduced reason for closeness (longitudinally) of stirrups is "confinement of the struts (in the analogy of Mörsch)". So we need to devise a feasible strut mechanism able to stand the loads. For loads atop (invert the position if otherwise), one of this can be a fan (an inverted triangle) taking loads from atop, then dividing in 2 struts, one towards each lateral leg of a stirrup. And the "tributary" load to be taken as the load passed in the truss analogy. In section projection the strut can be to some extent somewhat wide, since at bottom has a net of receiving strong longitudinal bars, and atop the node fan-struts. Contrarily looking from the sides the critical strut can't be taken as wide since loading in legs of stirrups too far apart.Say, 15 cm, 20 cm, 25 cm perhaps for the side width of the strut. And then you need to check for the strength of this tridimensionally confined strut. This will provide some extra strength but better not be too optimistic, it is supposed we are trying to check a thing under strut and tie provisions in the code. If the struts and the tie then work, the structure may work at the limit state and if the values you have used in the check are code-derived you have a handle to the code problem as well.

RE: capacity check - shear links

(OP)
just an update; i have got in front of me the handbook to the code. I will repeat the comment for the relevent clause below; Sorry for the lengthy post.

The logic behind the limit on the lateral spacing is less clear but experimental evidence suggests a reason why such a limit is valuable. One of the reasons of stirrup is to inhibit dowel failure of the tension steel....Clearly the effectiveness of the stirrup in achieving this will reduce with increasing distance of the vertical leg from the bar considered. Clearly, if a bar is placed further than 150mm from a stirrup leg, it can still be used to provide flexural strength but should be ignored in assessing Vc (for non UK this is concrete shear strength based on longitudinal reinforcement). The situation in slabs can be used to extend this interperation further. In slabs, stirrups are not required until V > Vc. It seems logical to argue from this that the requirement relating to the spacing of stirrup legs is to ensure that Vc can be maintained in circumatances where V > Vc. It therefore seems reasonable to conclude that the limitations on lateral spacing may be ignored where V < Vc.  

RE: capacity check - shear links

(OP)
now my own comment on my post above;

the code for beams requires links if 0.5Vc < V < Vc + 04 for minumum links and designed if greater value.

Slabs it is simply V > Vc.

So to me the above interpretation doesn't add up for the following reasons;

1)Beams and slabs are treated differently the last sentance seems to go against the code equations above
2) If we have to ignore longitudinal bars in beams that are not within 150mm of stirrup how can we use these bars in slabs? Surely if beams and slabs are comaprable you would have no longitudinal steel contributing to shear strength since most slabs are detailed without shear reinforcement.  

RE: capacity check - shear links

I have also some incongruence in design provided by one spanish structural design package, yet the cause may be (I think it is) EHE code itself. You have a slab on columns, check it, simply you deduce it is mandatory a thicker slab at support, whatever the rebar shear you add as per two way slabs design. Now, you include a cross of embedded beams (rebar shearhead) in the slab ... and complies without further thickening! It is clear then that the code must be more sympathetic to beam design than two way slabs for shear.

RE: capacity check - shear links

The Australian code, AS3600, is more consistent. (This may not help your problem but should help your understanding)

By AS3600, 0.5Vc doesn't apply for beams if D does not exceed the greater of 250mm and half the width of the web. The limit is Vc for those cases.

From the commentary to AS3600; "Concrete beams can possess onsiderable strength without shear reinforcement. However, this strength will be reduced if cracking occurs. Cracking resulting from restrained shrinkage and restrained thermal deformations have been responsible for a number of shear failures in members without shear reinforcement. Since shear failure can be quite sudden, the Standard adopts a conservative approach to the utilization of the strength of beams without shear reinforcement."

For some reason the same degree of conservatism wasn't necessary for slabs &shallow beams.

RE: capacity check - shear links

(OP)
can I have another opinion on this. How would you say the duration of loading would effect the shear failure? The load applied will be very short term. A one off crane lift.  

RE: capacity check - shear links

The process of loading through a cable of relatively small stiffness makes that the impact factor when applying the maximum load must be small, and even so more thant enough, it is my view, to overcome any loss of strength that would be required for long term applications. The assumed strength in the codes uses to contain around a 15% of diminished strength in compression for concrete for long term application of loads. So in general, being in shear a fragile failure, it is likely that if it survives short term would survive long term, with the crane load applied for some time, and normally not a very special other consideration of impact factor than as per recommendation of the crane operator needs to be accounted. If he says, here 35 tons reaction, put those there; most likely is a conservative estimate, both in value and since not truly long term, for shear.

RE: capacity check - shear links

(OP)
I'm sorry I am not understanding your reply. I am asking for opinions on the load duration for shear failure. Not whether there will be a long term problem even if it stays up during the crane lift.
The crane load given causes a problem as per the earlier posts.  

RE: capacity check - shear links

Essentially I am saying that you need not to use a particular impact factor on a sudden load, that ordinary calculations will cover for that once you use the calculated value of the reactions. Then as it is a temporary load, you may use the reduced safety factors allowed for works.

Conversely, I don't see a way of acknowledging a better behaviour for the concrete respect shear because the load lasts, say, 1 day or part of it. If the load brings the member to its (code) shear strength plus safety factor (the load safety factor consumed), you bring the structure at a situation that is not resisted every 95% of the times, or so goes the limit states theory.

So the safety factor on loading is what separates the beam of that degree of safety, to be safe 95% of the times. And this level of safety is sought to be attained anywhere reliably in the structure by following the normative clauses. So for longterm structures you keep 1.5 or 1.6 times the required strength to be a sound structure 95% of the times, and for temporary works, 1.25 or 1.35 times such required strength (less, but still well over the required strength to be safe 95% of the times).

Shear is a fragile failure and the mere average shear stress is a good indicator of its closeness to failure. For ordinary concretes of say 20 MPa compressive strength of the past decades, if you showed to have over 3.5 MPa in shear in a section, you were looking at rubble in a steel cage.

Nawy puts the limit of shear stress of plain concrete at about 20% of the compressive strength (without concomitant compression). Essentially, we do NOT want the plain concrete to crack in shear since a fragile condition, so we normally project our sections to limit values that forfeit before than anything the shear failure of the plain concrete within. We are when reinforcing in shear, in a situation akin to when reinforcing glass with some mesh, or some normal (quite deep) footing with a mesh. In all the three cases, it is expected that the plain material will crack before the steel does what it can; for the glass case most of the times just keep the pieces together; for the footing case, it is unlikely you will have more strength from the mesh rebar than of the bare concrete section itself but at significant amounts of rebar unusual till recently mandated at some places; and for the shear case, if you let to go the average stress in shear (without concomitant compression over the section) to 20% of the actual compressive strength, you will have a ruined member.

RE: capacity check - shear links

Duration of load is not a consideration in concrete design for shear.  Beam shear does not in fact occur instantaneously, but no allowance is made for short term loading as is done with wood.  

RE: capacity check - shear links

Perhaps a reduction in safety factor can be used, but a careful assessment & control of risks is required.

RE: capacity check - shear links

Handbook to British Standard by Palladian Publications suggests

1  The transverse spacing limit is basically to prevent dowel failure of the tension reinforcement
2  Any bar > 150mm from a stirrup leg should be ignored in the calculation of vc, but ccan still be included for flexure.
3  The spacing limit can be ignored if vu < vc.

RE: capacity check - shear links

(OP)
rapt, yes I am aware of this explanation. Folowing on would you then utilise the shear links (vertical bars) as resisting shear.

i.e. if you have your Vc based on the bars in 150mm of a leg, in my case the first two bars on the outside of the section. The Vu is greater than Vc can you include the links in the shear resistance? I only have vertical legs on the outside (side faces) of the beam. Probably explained better by my sketch attached in an earlier post.   

RE: capacity check - shear links

You can, but when in the shear calculation stating the longitudinal steel only count the first two bars, each side. Or at least, that is not in lack of logic or prohibited by the code. Furthermore, quite limited minimum longitudinal steel serves to ensure quite efficient behaviour of the stirrups.

But this can't make for the lack of compliance on transversal separation of vertical legs of shear reinforcement.

RE: capacity check - shear links

(OP)
yes I understand that interpretation of the code but I would then have Vc based on 4 bars (2 outer bars each side 150mm apart) in combination with 2 shear legs at these bars, so one vertical leg each side of the section. In this case The beam is 1500mm side so I have 1500 - 2 x 200 = 1100mm width of concrete in the middle with a) no shear links b) no effective longitudinal reinforcement from a shear resistance point of view.  

RE: capacity check - shear links

True. Can you define the section, what reinforcement above, what below, and shear links @ center ? this way we can make some assessment by other means, worksheets etc. I made a worksheet following a procedure by Vijaya Rangan that gave nicely (assume it works for your case) and adjusted minimum shear steel for a standing longitudinal steel (that defined at BS, tensile, above, etc) Also others following Collins etc that in order to get appraisal of the problem. On the other hand, I think for the proportions of your sketch and being a service level situation, the load most likely will arch towards existing (enough) stirrup steel at the sides...initially seems something more of pure code compliance than one actual problem at works ... as it used to be said, RC structures behave as we reinforce them. Here the designer has placed it seems enough shear steel but at faces, then structure will follow developing arcing action towards the stirrups and the only worrying things would be that in one strut and tie scheme compressive strengj exceed the limit and not enough tie be present. If the strut and tie scheme is solid, the structure will behave well.

RE: capacity check - shear links

I attach a printout of your case as an entry in the as per Vijaya Rangan model article worksheet. I have assumed an actual compression capacity of 20 MPa, so characteristic compressive strength should be about 30 MPa in the actual beam. For this data, smeared shear (not an area where strut and tie would make a better model) the worksheet indicates that your section could sustain some aesthetic factored 30 m·tonne moment in concurrence with factored 120 tonne shear. The struts neither would crash nor show cracking. The entry for the strength of concrete is one assumed to exist in the beam, so you can go from the actual strength of the concrete in the beam (85% of it as a precaution against sustained loads that fail structures 1 day after application).

A repeat of the worksheet with the same moment and fc at 10 MPa gives 85 tonne shear concurrent with 30 m·tonne (both factored), and with fc entry at 30 MPa the 30 m·tonne factored moment can be concurrent with 155 tonne factored shear. By tonne understand metric ton. It seems all quite linear growth in bearable shear with strength. In all three fc cases, at separation 15 cm you are over twice closer than required as a minimum (as the sheet was made). In all cases shear cracking and strut compression remains safely apart from limit values, so a failure of the strut and tie scheme seems unlikely.

Please look the worksheet and look specially for some area of bad input (input goes in blue), gladly may approach more another copy to your case.  

RE: capacity check - shear links

The case for fc= 10 MPa wouldn't be covered, really, see the margins o applicability. But from 20 MPa would apply to the actual strength of almost any practical modern structural concrete since the eighties.

RE: capacity check - shear links

Whilst in a walk I realize that what above I have called a "cosmetic" moment it is not so for the reduced number of longitudinal bars being accounted. The program is for shear field action and then needs vertical and horizontal reinforcement available to sustain the "strut" action hence a severe reduction on steel available for this action reflects inmediately on the shear available. That means that if you consume all longitudinal steel for moment, the model can't return much above plain concrete shear strength, so don't ring the bells for now. Soon after using a bit more the worksheet will post what found.

RE: capacity check - shear links

For the moment disregard the printout given above for the moment was excessive for the allowed longitudinal steel.

RE: capacity check - shear links

The worksheet is for optimization and showing scarcely sensitive to the change in longitudinal steel. Have other variants of the same that could be, but rather try another approach with others about shear to see some cooked solution within my tools.

 

RE: capacity check - shear links

It would be useful to have Mu Vu concurrent for design. For the moment, accounting with the 9 16 mm diameter bars at one side, if crack control is to be respected, the reinforcement would be allowed to take a bit less than 120 metric tons in shear by the spanish EHE-98 (former) code ruling RC. If you take 5 bars out of these, this is reduced to around 85 tons (factored). This is for fck=30 MPa that would allow for fcd=20 MPa.

Estimates according to Rahal give even lower than by the previous evaluation.

One evaluation (on Gaetano@Puleri at ACI SJ) goes to 480 metric tons for fc=30 (without fi reduction). Maybe not as dissimilar as apparent, they surely are targeting the actual final value (put, as quoted above, by Nawy, about 20% of fc, here shown less percentage, depth effect likely) and so not comparable to code practice. As much difference we must expect between what we design and the limit strength, 3.5 for ordinary checks may be quite likely and over 4 for connections and risky checks, like fragility in shear. This is not a code check, but makes clear that one shouldn't expect THIS member fail quickly in shear.

A check as per ACI as I have it in a render of it would accept 190 tonne factored shear.

So you have a whole gamut of opinions about your member.

Hope this helps.

 

RE: capacity check - shear links

Herewegothen. For the reinforcment layout, did you pull these from a drawing? Did you field verify if there were additional links installed (maybe a construction change that wasn't documented). Maybe some non-destructive testing is in order.

On another note, isn't there an exception in the code that allows no stirrups to be installed if Vu < 0.5*phi*Vc (See ACI 318 11.5.6.1). Also in ACI 318 sec 7.11, you should pay close attention to the wording which activates the tie spacing requirements for "compression reinforcement in beams." If you aren't using the longtudinal bars for compression, then you don't need to meet thetie spacing requirments.

If all else fails, propose strengthening (vertical epoxy dowels, fiber wrapping, ect...).

RE: capacity check - shear links

Also, if a shear field was of order, the skin reinforcement would be counted.

RE: capacity check - shear links

(OP)
in depth, no we have not feild verified but I have  a couple of drawings showing the detail, one from the client one from the original designer. Also the calculated shear stress requires designed links i.e. is above the lmit you show to leave out links. I believe I have a shear approching 1200kN from memory.

Field chack is the next step, the structure is only rented so any remedial work opens up a new can of worms

RE: capacity check - shear links

Here go 3 worksheets, one that of Vijaya Rangan method used for optimization, assuming the 4 16 mm diameter bars are available for shear-field action. The two others follow Collins&Mitchell estimate for shear in one of their books. The first would allow for Vd=1270 kN, and the two others may rate the concrete and stirrups contribution at 980 kN and 1448 kN, hence making a shear load at 1200 kN an unlikely problem.

I recently downloaded the freely downloadable RESPONSE series of programs and may use your case to try its use.

RE: capacity check - shear links

I installed my RESPONSE 2000 copy (a program by Evan C. Bentz at the University of Toronto, directed by Michael P. Collins) and ran your problem. You can download a series of 4 programs, Response 2000 one of them at

http://www.ecf.utoronto.ca/~bentz/download.htm

In the first page note your section (forfeit the loads, these are not those later applied). Note that a 15M bar is accepted to have 200 mm2 section at the program.

RE: capacity check - shear links

And then see how to 226 kN·m behaviour goes nicely.

In all, even when the program is for shear-field action and so not the best choice where strut and tie schemes are, note that by this MORE realistic assumption than a pure Mörsch scheme there is still no cracks at your section, and the implied deformation has not even started to pass stress to the stirrup. So this is another favorable opinion on that the section will behave well.

You see in this model I have counted the 9 bars atop and at bottom, and not included any skin steel. I plan to enter the 4 above 4 bottom plus some skin steel case later; if something negative develops I will let you know, otherwise give this free nice program a try, it takes 15 mins to get these results.

RE: capacity check - shear links

(OP)
thanks very much, Ill look into it

RE: capacity check - shear links

(OP)
Update. I have inspected the structure. There is no visible cracking of the effected beams. Of course I do not know the actual loads applied to the structure or if they are any where near the design loads but at least I haven't seen any visible cracking due to shear.
I am awaiting further loading information in order to proceed.  

RE: capacity check - shear links

herewegothen,

A few thoughts here - I think what hokie66 suggested above might be the direction I'd take - look at a reduced Vc based on what you DO have in terms of longitudinal bars contributing to shear strength.  

In the US (ACI 318) I don't believe there is any provision similar to yours where stirrups must be spaced out across tension bars.  Compression bars yes.  I've always added additional links(stirrups) for wider beams though...just for feel good reasons.

The shear provisions for slabs vs. beams is based upon the idea that slabs can transfer load across slab widths and thus add to some redundancy.  Beams are single, non-redundant elements so more conservatism required.

One "fix" idea:  Could you drill vertically through the beam and install vertical bars in grout/epoxy/adhesive?...or perhaps plates top and bottom?  Might be a lot of work and you might have struggles with hitting the longitudinal bars but since you said you were "stuck" this fix came to mind.

 

RE: capacity check - shear links

Did I say that?

RE: capacity check - shear links

hokie66 - I must have been smoking something - I think I took your statement:

"If it is a wide, flat beam, you should be able to use the one way slab provisions rather than the beam provisions"

as using slab provision vs. beam provisions and maybe getting it to work....I was probably half asleep too.

RE: capacity check - shear links

I think ACI has questioned the effectiveness of a beam  wider than the column. Just from memory, check it out.

RE: capacity check - shear links

Huh?  More often than not, beams are wider than columns, seismic requirements not controlling.  They shouldn't be the same width, as it makes the reinforcement clash.

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