Designing Columns for Concrete Buildings
Designing Columns for Concrete Buildings
(OP)
I'm designing a 15 story post tensioned building and I'm trying to design the columns in Risa Floor. I'm just wondering what the standard practice is for considering moments from the slab in the design of the column.
1) I want to design the columns as axial compression members only but I know the slab will put moment into the columns. Is this a standard assumption that I am required to make and design for?
2) When I model the floor I connect all the columns with fixed-fixed beams so that the columns will suck in moment in the Finite Element analysis.
- Will this put too much moment into the columns? I have a 3 span condition 13'-27'-13' so theres a lot of unbalanced moment from the dead load alone.
4) If what I'm doing is overkill for the moment, do you model the slab moment connection as spring, and how?
I have slab cantilevers too. Does anyone have the magic bullet for designing these quickly. Thanks!
1) I want to design the columns as axial compression members only but I know the slab will put moment into the columns. Is this a standard assumption that I am required to make and design for?
2) When I model the floor I connect all the columns with fixed-fixed beams so that the columns will suck in moment in the Finite Element analysis.
- Will this put too much moment into the columns? I have a 3 span condition 13'-27'-13' so theres a lot of unbalanced moment from the dead load alone.
4) If what I'm doing is overkill for the moment, do you model the slab moment connection as spring, and how?
I have slab cantilevers too. Does anyone have the magic bullet for designing these quickly. Thanks!






RE: Designing Columns for Concrete Buildings
1) Yes, the standard assumption is that you have to design the columns for all load effects on them. This includes vertical load effects and horizontal load effects, which should be significant in 1 15 storey concrete frame building, resulting in axial forces and moments on each column. You cannot design them as compression only!
2) I am not sure what fixed-fixed beams are. I hope you are only modelling the fixity that actually exists and can be generated by the frame!
You cannot deliberately over stiffen the connection to make the slab design work better. You should be analysing the frame as it will act in practice. Some designers reduce column stiffness for the frame design (in Asia they quite often ignore the columns completely in the slab/beam analysis) but that is dangerous as the column will attract moment and failures such as punching shear must be checked based on the real frame stiffnesses as redistribution is not possible. Columns in this case still need to be designed for the loads and moments that they attract.
Why do cantilevers make it difficult. I do not understand the NEED to design these quickly. You need to design them properly!
I hope you are doing this design under the supervision of an experienced concrete building designer!
RE: Designing Columns for Concrete Buildings
There are many ways to estimate this stiffness for columns, because of their compression bending interaction they are not the same as beams and slabs, thus you need to do a bit more home work. for a first run situation I normally use 0.7Ig as a starting point to size your columns then use have to use a empirical approach, there was a good discussion in a paper in the July 2009 ACI journal, by Kenneth. There is a good publication by Rice and hoffman that could help as well.
As for cantilevers these are something that you want to spend time on getting the stiffness correct and making sure they are more than strong enough as they have no redundant system.
When in doubt, just take the next small step.
RE: Designing Columns for Concrete Buildings
1) Always consider the moment that is being transferred into the column from the slab. Also because the moment capacity of the column is dependent on how much compressive stress the column is subject to, look at a maximum and minimum compressive load envelope. The greatest moment in the column may occur for a patterned or checkered live load case which will reduce the axial stress and may have a reduced capacity depending on where the load points lie on the interaction diagram.
2) Are you referring to fixed-ended columns at the floor levels above and below the floor you are designing? Or are you referring to fixed-end moments used in frame design by moment distribution/redistribution?
3) To be on the conservative side, when designing columns design them with full stiffness. i.e. Icol=Igross. Some codes allow for the analysis to be conducted using 0.8*Ic. This allows for some stress to be redistributed because of the loss in stiffness that occurs when a section cracks.
Cantilevers can be designed quite easily, they are determinate structures, the moment cannot be redistributed. A spreadsheet can be set-up quite quickly to design overhangs.
Does risa analyse equivalent frames or meshes the slab and designs by FE?
RE: Designing Columns for Concrete Buildings
RE: Designing Columns for Concrete Buildings
There is no easy button for multi-story concrete columns but I've found Etabs does a nice job. Although I don't always agree with how it determines sway vs non-sway. Still a lengthy, input and multiple runs process to economize and back check.
RE: Designing Columns for Concrete Buildings
In a 15 storey building, under vertical loads, the columns at the lower floors will not crack as the axial compression effect will be far greater than the moment effect and the column will be in compression. Why are you using .7Ig if the columns cannot crack. You can only allow for reduced stiffness if there is going to be cracking.
The article you are talking about in ACI is discussing stiffness for sway cases as I understand it. You cannot use this logic for vertical effects.
RE: Designing Columns for Concrete Buildings
CRAPT,
I want to design them quickly because this is just a preliminary design for estimating purposes for a design-build project. I need to get column sizes to the architect so they can lay out their guest rooms. In Risa the beams are fully fixed to the columns. I modeled the beams as the depth of the slab x the tributary width wide, then the program will calculate the correct Ig of the beams which will produce the correct stiffness in the equivalent frame analysis. The building has a high aspect ratio so I figured that the critical direction to analyze the columns was in the short direction where the column spacing's are not the same, so I ran the floor slabs as one way so that the load went directly to the beams first and then to the columns. This is how the equivalent frames in the short direction are analyzed anyway so I figure that I can pull the end moments from the beams out of Risa and use these for the Post Tension slab ultimate strength design.
ROWENGINEER,
I'm using the full Ig for the column stiffness. I think this is conservative for the column design because this pulls the maximum moment out of slab and into the column. But for the slab design I would think that you do the opposite -> use the weakest column (0.7Ig) or maybe even no column stiffness so that all the moments stay in the slab so that there is adequate steel in the slab.
ASixth,
I'm referring to fix ended beams. There is no FE plate in the Risa slab, I'm hoping that these beams I'm modeling serve as column strips to collect the load and have a stiffness for the equivalent frame analysis. I have a spreadsheet now that I'm using for the PT slab design (since there are so many stress checks) and I'm pulling the end moments from these beam strips out of Risa to design for. For the lateral loads I might just model an entire 15 story frame and push the frame 3" to the right (h/500) and see what moments develop. What load case would I use for this? 1.2+1.6(moments developed from displacement)?
Spats,
To design for creep, you multiply the deflections by the factor in ACI right? To design for temperature, we limit the length of the building? TO design for shrinkage, we limit the length of the pours and use pour strips? But to account for the PT forces in the columns, aren't these good for column? Don't they relieve the dead load moments that the equivalent frame calculates? Because isn't the PT force effectively a negative load that counteracts the dead load?
CTSENG,
If I prove that these gravity columns are non- sway, then do I have to analyze them for lateral loads?
RE: Designing Columns for Concrete Buildings
Yes, you still must include all applicable load combinations in the design of the columns. The sway vs. non-sway criteria affects how you magnify the column moments.
Now in a non-sway frame, moments from the lateral forces might be low, but that depends upon the way the structure is braced, how combinations of alternating live loads and lateral loads interact, etc. But your lateral forces will possibly produce some moments in the columns and they would simply be included in the M2 values of the non-sway derivation of the δns values in ACI section 10.12.3.
RE: Designing Columns for Concrete Buildings
You keep referring to Fixed Ended beams. That is mixing us up. A fixed ended beam has full fixity, which you do not have. I assume your beams have a full moment connection to the column and the frame analysis sorts out where the moments go. If they were fixed ended, the moment would be wl^2/12 at each end and wl^2/24 at mid span and would be very unconservative.
Agree with JAE that you need to allow for sway moments no matter what the framing system. In a braced building there is still sway and some sway moments and axial forces go into the columns.
You have to look at all of the possible combinations of wind, earthquake, and vertical load effects that the your loading/design code requires.
For the sway case, it is usual to use reduced column and beam stiffnesses as the members will usually be cracked and will loose stiffness, resulting in larger sway effects and more cracking.
PS why should you get another career?
RE: Designing Columns for Concrete Buildings
Creep, shrinkage, temerature & post-tensioning forces all represent VOLUME CHANGES in the structure. Volume changes can significantly affect the stresses in your concrete frames... more so with taller buildings. Temperature is more important in exposed structures such as a CIP parking garage than it is in your instance. Temerature differences between the upper and lower surface of a beam/slab are very important when analyzing a garage, as are changes in member lengths.
You design for volume changes by including them as a load condition in your analysis program. I'm not familiar with RISA. I do this type of analysis using STAAD. With STAAD, you can apply temperature loads, pre/post stress loads, as well as strain loads (creep & shrinkage).
To obtain guideance on what kind of strains to figure for creep & shrinkage, I would reference ACI publications, and the ACI 209 Committee.
RE: Designing Columns for Concrete Buildings
I wanted to comment on your statement:
"But to account for the PT forces in the columns, aren't these good for column? Don't they relieve the dead load moments that the equivalent frame calculates? Because isn't the PT force effectively a negative load that counteracts the dead load?"
The PT "lift" really doesn't help the columns. The total load to be resisted by all of the columns in floor should remain the same. At best you might simply end up shifting some load from one column to another.
When you stress a PT floor slab, it tends to shrink axially. The specifics of this depend on tendon layout, shear wall distribution, etc. This shrinkage can result in large shear loads being applied to the tops of your columns as they go along for the ride. I believe that this is what Spats was referring to.
A colleague of mine designed a building similar to yours (mid rise, PT slab as lateral frame). We talked about it often. In retrospect, I'm not sure that I care for the system. Some of the reasons why:
1) An efficient PT slab is pretty thin. It doesn't do a whole lot for the frame stiffness. Especially when you consider the equivalent column business, which you should. Yeah, I realize that the prestressing stiffens the slab some.
2) We had no drop panels. As such the moments caused by lateral loads really exacerbated our punching shear problems. In general, I question the ability of these joints to accomplish the shear & moment transfer required of them when used in frames.
3) There isn't all that much mild reinforcement in a typical PT floor slab. In my mind, this seriously limits ductility and opportunities for load redistribution.
I try to steer PT floor slab buildings towards using shear walls for lateral resistance. In a high seismic area, that's the only way that I'd use PT floors.
KK
RE: Designing Columns for Concrete Buildings
Thanks for the pickup.
I use 0.7Ig for my first run in Finite element modelling, just so I can have one model for the first estimate PRELIMINARY DESIGN (assumption was due to the need for quick and dirty this was a preliminary sizing design, not as it turns out a complete reo design). I wouldn't normally use a finite element package to perform this task myself unless the building is complex and has a few transfer floors, with cokmplex column locations. I have noticed however that a lot of colleges, whom are heavy finite element users, prefer to do a complete model for preliminary design.
The reason for the 0.7Ig is for horizontal loads, I had assumed (probably wrong) this being a PT building, with thin slabs (180) that this was going to a quick column slab operation building with a small core. Here you want the columns to do a bit of work for you with regards to horizontal loads normally all I care about in preliminary design is getting serviceability criteria correct, because undersize members in relation to service requirements are very hard to fix during detailed design, however fixing strength problems is relatively easy.
However as the OP is doing a reo design no sizing design, I would suggest that he ignore my earlier advice, which he seems to be doing.
When in doubt, just take the next small step.
RE: Designing Columns for Concrete Buildings
In simple if you are in SDC A, B, C you can design your columns for gravity only, please do check the moments cranked in by slab. In most cases if you provide 1.5-2% reinf. you will be ok.
RE: Designing Columns for Concrete Buildings
I'm checking the punching shear of the slab now and there are incredably high moments getting sucked into the columns from the slab. This is killing the punching shear because of the unbalanced moments. I went back and analyzed the unbalanced moments again by considering that the PT force releives these unbalanced moments so in my model of the equivalent frame I simply superimposed the wb (effective uplift from PT) to the LL and DL in order to get unbalanced moment. Is this ok?
Rapt, Rowingengineer,
Sorry for the confusion, Fixed-Fixed beams refer to that the element is fully fixed at both ends in the model and the FE analysis sorts out where the moments go. A question reguarding Ig and Ie; aren't these close to the same? I always figured that if you use the min required flexure reinforcement 3 sqrt(f'c)bwd/fy then they should be close to the same since this requirement ensures the same ductility and strain compatability as a linear gross concrete cross section. If anything Ie should be higher. This is a separate study. I should get another career because this is rediculous.
StrucEng,
Thanks for the reference that will save me.
General,
In order to get the forces to design the slab and columns, will this procedure work (The procedure I have already implemented)...
1) I modeled a full 15 story section of the equivalent frame. 4 Column lines and 3 spans x 15 stories
2) Column sizes are as needed and beams between columns are slab thickness x L2 so that the propper Ig is used for stiffness.
3) If I run the DL+LL case can I use these moments to design the slab? + Reduced to face of support
4) To design the columns can I reduce the proper live load of each floor (Typically 40%LL for all floors) and use these axial forces and moments to design the columns.
5) Would it be wrong to reduce these moment further by allowing the wb (dead load balancing moments from PT)to relieve the moments being sucked into the column?
6) To get the unbalanced moments for the punching shear checks I'm simply subtracting the end moments from adjacent beams (which should give me the moment going into the column). Is this what the unbalanced moment is? And will thsi work? Can I let the PT balancing forces to releive these moments like I suggested in 5?
RE: Designing Columns for Concrete Buildings
It still seems like the PT forces will help the columns out.
RE: Designing Columns for Concrete Buildings
You can try to rationalize away volume changes if you like. It only makes you terribly wrong.
RE: Designing Columns for Concrete Buildings
You're right, the PT balancing forces will definitely help the columns out with respect to moment and therefore punching shear as well. The point that I was trying to make is that the PT won't significantly alter column axial loads. Sorry if I confused matters with my comment.
KK
RE: Designing Columns for Concrete Buildings
Arguing with an engineer is like wrestling with a pig in mud. After a while you realize that them like it
RE: Designing Columns for Concrete Buildings
To rephrase: I'm trying to say that Ig should approximately equal Ieff.
RE: Designing Columns for Concrete Buildings
The only PT effect put into the columns for ultimate design is the secondary effects. You cannot reduce the loading by the balanced load as the PT puts a vertical force into the column as well as uplift forces between columns, giving a nett uplift effect of ZERO (gravity still exists and PT is not a skyhook!
The total vertical force going into the columns in DL + SDL + LL all factored + PT Secondary.
PT secondary reaction for the whole floor is ZERO.
So the vertical loads in the columns are the same with/without PT except for some very minor redistribution between columns. Yes, the moments will normally be reduced by the secondary prestress moments, which are significant for end columns.
RE restraint effects, they are maximum for the bottom couple of floors and the roof. They can be much small for the floors between, depending on the stiffness of the end columns or if there are multiple cores. You cannot make a general statement on this.
RE: Designing Columns for Concrete Buildings
Arguing with an engineer is like wrestling with a pig in mud. After a while you realize that them like it
RE: Designing Columns for Concrete Buildings
RE: Designing Columns for Concrete Buildings
Concerning your statement:
"RE restraint effects, they are maximum for the bottom couple of floors and the roof. They can be much small for the floors between, depending on the stiffness of the end columns or if there are multiple cores. You cannot make a general statement on this."
I haven't heard this before. For the lower floors, is it because of the proximity to the foundations and lateral support at grade? What's the rationale for the roof?
Thanks,
KK
RE: Designing Columns for Concrete Buildings
I should have split the restraint effects up a bit.
For shrinkage and temperature shortening and PT shortening, the lower floors are affected badly because there is no movement possible at the foundations. The stiffer the edge supports, the higher up the building this goes. If you have end shear walls, it could be a reasonable height up the building. You would have to analyse the full building for a shrinkage effect to find out hoiwe far up the building is affected.
For PT shortening , the roof is also a problem, depending again on end support stiffness. For lower floors, when you stress at a floor, restraint from the columns below reduces the amount of PT axial force that goes into that floor. But the restraint is caused by the the columns reacting against the floor below, so the axial force that is not going into the current floor gors into the floor below (except for its restraint again), making up for much of its own loss when it was stressed.
For the roof, there is no floor above to be stressed, so it does not get back some of its lost force as the other floors do.
The bottom floor does not get much back because the foundation restraint is so large.
RE: Designing Columns for Concrete Buildings
The secondary reactions should also be included! The secondary moments are caused by column reactions to the prestressing!
RE: Designing Columns for Concrete Buildings
If you are going to design saying that Ieff=Ig (or that Ieff/Ig=1) then you should check to ensure that the section never exceeds it's cracking moment. Every research report or technical document that I have ever read all show a graph plotting deflections versus moment. They all exhibit linear behavior up to the cracking moment and then display non-linear deformation behavior post-cracking.
I make this assumption (Ieff=Ig) when designing prestressed girders (using high performance concrete as you know) but I ensure that at every point along the girder the cracking moment is never exceeded.
Attached is another paper for your collection.
RE: Designing Columns for Concrete Buildings
What secondary reactions are on the columns? I'm just assuming moment reactions (M) from eccentrity of the tendons. Don't most of the Fx (horizontal) reactions get absorbed by the slab? and Fy (vertical) reactions are not existant as they disobey the laws of physics (If they push the column down then there must be an opposite and equal force pushing the building up.
Asixth,
Ig is the moment of inertia of the section with only concrete and no steel considered. I think what you meen is that Ieff=I ("I" being just "I" and only "I" with no subscipts because it is the true moment of inertia of the section considering the steel). I'm arguing that Ig is aproximately Ieff for most ranges of cracking. When I visualize a rectangular cross section of just concrete and then juxtupose the same size rectangular section minus the cracked portion plus transformed steel, then the section I feel would have a similiar moment of inertia.
RE: Designing Columns for Concrete Buildings
I think you need to put some real numbers to your Ief statement.
The Australian code has a couple of simplified formulas that may help you get a better feel for the large reduction in stiffness.
For reinforced rectangular sections
Ief = (0.02 + 2.5p)bd3 where p < 0.005 (percentage of tensile reinf).
&
Ief = (0.1-13.5p)bd3 where p > 0.005
This results in (approx):
(p = 0.5%) Ig = 0.39 Ig
(p = 0.75%) Ig = 0.465 Ig
(p = 1.0%) Ig = 0.54 Ig
(p = 1.5%) Ig = 0.69 Ig
These are obviously conservative (deemed to comply) but you can see the trend.
Hope this helps.
Rapt - your use of columns to tension lower stories of buildings is impressive! Might be able to utilise this principal to upgrade existing buildings! What would happen at a 'soft story' ie discontinuous shear walls etc..? Haven't got my head half way around all the restraint issues!
RE: Designing Columns for Concrete Buildings
Those cracked figures are for flexural members, so no axial compression has been allowed for. The Ief figures would be a lot larger in a column where there is axial compression, and in a 15 storey building, at the lower floors the columns would be in compression under vertical loads, so no cracking. and Ief = Ig.
RE the prestress effect when stressing, it is simple statics.
Actually, the upper floor prestress compresses the floor below if there is restraint from the supports below the floor being stressed!
RE: Designing Columns for Concrete Buildings
We are talking about Prestress Secondary moments aren't we! I assume you know the difference between the full prestress moment (Mp below) and the prestress secondary moments (Msec below)! The prestress moment is not just the eccentricity moment P * eccentricity!
Mp = P * eccentricity + Msec
Wherever there is a moment in a frame, there is a shear and therefore a reaction.
The Msec moments are actually caused by column reactions stopping the slab lifting off the columns under the prestress forces because they are connected to the columns.
The secondary moment case puts moments and reactions into the columns. The total of the prestress reactions for a member or a floor is zero so there is no net vertical force generated, but there is a rearrangemenmt of column reactions due to these forces.
So each column gets a Mp and an Rp from the prestress which needs to be included in the column forces. Note that Rp is normally realitively small compared to the other reactions.
RE: Designing Columns for Concrete Buildings
Note we are discussing a post tensioned building,what the P/A is I have not idea, but the Ig for a Pt slab after cracking is different to a reinforced beam.
I would never design a pre-stressed slab to crack under service loads, would be counter productive. also after cracking the PT slab behaves in a completely different manner compared to a reinforced slab. The restraint and temp effects may cause some cracking and a reduction in Ieff (I like to use about 0.7Ieff for slabs), but i will let other argue that point.
If you are however concerned that you may get different moment at ultimate state due to cracking of the concrete, generally this would only be of consequence for torsion beams thus i wouldn't get all that worried for a beam and slab arrangement due to cracking of the concrete.
asixth,
Thanks for the paper, but his book is a better read.
Arguing with an engineer is like wrestling with a pig in mud. After a while you realize that them like it
RE: Designing Columns for Concrete Buildings
RE: Designing Columns for Concrete Buildings
Arguing with an engineer is like wrestling with a pig in mud. After a while you realize that them like it
RE: Designing Columns for Concrete Buildings
"I would never design a pre-stressed slab to crack under service loads, would be counter productive. also after cracking the PT slab behaves in a completely different manner compared to a reinforced slab. The restraint and temp effects may cause some cracking and a reduction in Ieff (I like to use about 0.7Ieff for slabs), but i will let other argue that point."
Most PT slabs designed in Australia over the last 35 years have been designed to crack, at least at the critical sections. It is not economical to design PT slabs as uncracked and their performance is far better than RC slabs.
How do they act in a different manner to RC slabs after cracking and why is it a problem?
Yes, Time-effects in Concrete Structures is a great book, but I thought it was out of print!
RE: Designing Columns for Concrete Buildings
no problems with the way concrete cracks I was just pointing out that the %Ig value given by OzEng80 are not applicable to post tension slabs.
"Time effects in concrete" is ou of print is news to me, but i did buy mine a while back.
Arguing with an engineer is like wrestling with a pig in mud. After a while you realize that them like it
RE: Designing Columns for Concrete Buildings
Concrete Structures - Stresses & Deformation
Mamdouh M. El-Badry
I had the good fortune to take a class from Dr. El-Badry. The methods presented in his book represent the state of the art and are a considerable improvement upon the methods traditionally used.
I don't mean to take anything away from Gilbert's work of course. I revere the classics as much as anybody.